SEATTLE'S SEISMIC RETROFIT PROGRAM David W. Korpi1 ...

SEATTLE’S SEISMIC RETROFIT PROGRAM
David W. Korpi 1 ; Robert Richardson 2 ; Yuling Teo 3
Abstract
The City of Seattle’s seismic retrofit program has been continued to address the
seismic vulnerabilities of seven additional bridges. This is being done as a result of the
City’s Bridging the Gap funding initiative. An Expert Review Panel of nationally
recognized seismic experts has been formed to review the seismic criteria, policies, and
the analysis work being performed. The latest FHWA manual is being used, but the City
has also tailored its own retrofit policy to best utilize its limited funds. Geotechnical
hazards of high seismicity and soil liquefaction make seismic retrofit expedient. The
South Albro Place Bridge seismic analysis and retrofit is highlighted.
Program Description
The City of Seattle 4 has developed an ongoing seismic retrofit program for the
approximately 120-plus bridges in their inventory. Beginning in the early 1990s, Phase I
of the retrofit program was implemented, and 23 bridges were retrofitted to various levels
of seismic performance. Now Phase II of this program is being implemented and seven
additional bridges are being retrofitted.
Seattle’s inventory of bridges is quite diverse in terms of bridge type, size,
complexity, date constructed, materials of construction, and current condition. This
variability is a result of local topography and water features, and layout of transportation
infrastructure for the City. In the 1990’s several of these major corridors received seismic
retrofits, although not all structures comprising any given corridor were upgraded to the
same level of performance. Funding was not available to complete all the retrofits that
were identified as necessary or desirable.
Soon after the 1989 Loma Prieta earthquake in San Francisco, the City of Seattle
approved funding for seismic retrofitting its bridges as a proactive measure in protecting
public safety and reducing the socio-economic impact to the community in the event of
an earthquake. This measure resulted in the birth of the City’s Seismic Retrofit Program
(Phase I) which was able to retrofit 23 bridges to be more resistant in the event of an
earthquake. Due to limited funding, the Program has been relatively inactive since the
completion of the first phase of retrofits.
1 Senior Project Manager, Bridge and Structures, HDR Engineering, Bellevue, WA
2 Project Manager, Bridge and Structures, HDR Engineering, Bellevue, WA
3 Supervising Structural Engineer, Seattle Department of Transportation, Seattle, WA
4 Hereinafter “Seattle”, “City”, and “SDOT” are synonymous.
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In 2006, Seattle voters passed a nine-year, $365 million levy for transportation
maintenance and improvements known as Bridging the Gap. The levy was complemented
by a commercial parking tax ($127.5 million) and an employee hour’s tax ($51.5
million). Over life of the levy the total expected revenue from the three sources is $544
million. In the nine-year period which began January 2007, the Bridging the Gap funds
will be available for transportation capital projects and the much needed infrastructure
maintenance. This work to be funded through Bridging the Gap includes the following:
• Paving and repairing streets
• Constructing and improving sidewalks
• Rehabilitating roadway structures including bridges, retaining walls, and
stairways
• Bridge seismic retrofitting
• Improving major corridors
As a result of this renewed emphasis and Bridging the Gap funding, the City
decided to reactivate the Seismic Retrofit Program, and the following bridges were
designated for Phase II:
1. Fauntleroy Expressway
2. Ballard Bridge
3. King St. Station Vicinity Bridges (includes four structures)
4. South Albro Place
Expert Review Panel
One of the successes of Phase I of the retrofit program was the utilization of a
review panel of noted seismic experts. They reviewed the criteria, seismic analysis,
strategies, and proposed retrofits. Phase I of the retrofit program was based upon the
FHWA 1995 manual (Ref. 5). Since this approach was relatively new and untested, the
City determined it was expedient to have the panel of experts involved in the Phase I of
the program from the beginning.
For the same reasons, and because it was so successful in Phase I, the City has
elected to gather a panel of experts to review the Phase II program work. For a list of
seismic experts on the Expert Review Panel (ERP), see the Acknowledgements.
Specific bridge seismic retrofit design criteria were developed for the program
based upon the bridge seismic retrofit philosophies and policies established by the City.
These criteria encompass current state of the art technology on the seismic retrofit of
bridges, and have been reviewed by the Expert Review Panel (ERP). The Bridge Seismic
Retrofit Criteria addressed technical, financial, functional, physical, and community
parameters/considerations.
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Seismic Retrofit Philosophy
The philosophy of the SDOT Bridge Seismic Retrofit Program is comprised of
two guiding principles:
1) First and foremost, provide for public safety in the event of an earthquake
2) Reduce the socio-economic impact of an earthquake to the community to a
reasonable extent
This philosophy led to the establishment of appropriate retrofit criteria. In keeping
with the philosophy for new bridge design (Ref. 2), the following general performance
objectives of the retrofit program were identified:
• Small to moderate earthquakes should be resisted within the elastic range,
without significant damage
• Realistic seismic ground motion intensities and forces should be used in the
design procedures
• Exposure to shaking from large earthquakes should not cause collapse of all or
part of the bridge. Where possible, damage that does occur should be readily
detectable and accessible for inspection and repair.
SDOT Retrofit Policies
The most recent seismic retrofit guidelines published by FHWA and MCEER
(Ref. 3) are based on a two-level approach, whereby for standard bridges operational
performance is desired for a smaller, more frequent earthquake and life-safety/no collapse
performance is desired in a larger event. While this is ideal, limited public funds has led
the City to make prevention of loss of life the paramount concern over bridge operations.
The reason for this is to improve the greatest portion of the bridge inventory with the
monies available. This observation is noted in the new LRFD provisions for seismic
design (Ref. 1), including the Guide Specification for new design (Ref. 2), where a single
level earthquake is used for design of new structures; no service level earthquake is
mandatory for consideration. Thus, a lower service level of earthquake was evaluated, but
based on costs, retrofitting for a lower level earthquake was considered optional.
The following policies have been adopted by the City for their retrofit program:
• Bridge Importance - All City bridges will be considered “Standard”.
• Earthquake Hazard Levels and Performance - A 1000-year return period (Upper
Level Earthquake) was used to design seismic retrofits to prevent collapse and
threats to life safety. A 100-year return period (Lower Level Earthquake) was
used to design retrofits to maintain operations after more frequent events.
• Remaining Service Life - A bridge that has 15 years or less of remaining service
life, or is scheduled for replacement within the next 15 years, will be considered
by SDOT on a case-by-case basis for retrofit. Each bridge will be considered by
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SDOT based on its own vulnerabilities, importance and likelihood of future
replacement. Additionally, if inexpensive retrofit measures, for example
restrainers, provide significant enhancement to life safety, then such retrofits may
be considered.
• Age of Bridge - Bridges built after 1983 and were designed with the AASHTO
1983 Guide Specification or later versions will not be considered as high priorities
for retrofit.
• Bridge Location / Usage - Small neighborhood bridges or pedestrian bridges over
minor, non-lifeline roadways will not be retrofitted.
• Holistic Retrofit Approach - The City will not adopt the phased approach to
retrofitting but rather will develop a complete list of retrofit needs for a given
bridge.
• Performance-Based Retrofit - The City will adopt a performance-based seismic
retrofit approach for its bridges. To the extent possible, more advanced
displacement-based assessment techniques, such as pushover (nonlinear static)
analysis, should be used as a minimum.
• Seismic Retrofits For Bridge Preservation - Seismic retrofit is viewed as a bridge
preservation strategy by SDOT, and not a general transportation improvement
strategy; thus, under this Program only the bridge features that require retrofitting
to satisfy seismic performance objectives will be constructed.
• Bridges Previously Retrofitted - Bridges that were retrofitted in previous phases
to either operational or life safety levels, and for which all identified retrofits were
constructed, will not be retrofitted further.
• Subsurface Foundation Retrofit - Investing in subsurface retrofit will be a low
priority unless the life-safety performance level cannot be met without such
retrofit.
• Retrofit vs. Replacement Cost Considerations - If cost of retrofit is a substantial
fraction of the replacement cost for a given bridge, SDOT may choose not to
retrofit and instead program the bridge for replacement.
• Liquefaction - The potential for liquefaction and liquefaction-induced hazards,
such as lateral spreading, will be identified in the geotechnical report. The effect
of liquefaction on lateral resistance and vertical support of the structure will be
considered.
Geotechnical Issues
High Seismicity
Seattle’s geology and geomorphology are as varied as its bridge inventory.
Geotechnical parameters range from sandstone to soft, lake clay with peat deposits. In
some instances soil characteristics vary significantly from one end of a bridge to the
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other. This requires a greater modeling complexity in order to determine appropriate soil
response under dynamic loading. In most cases, we utilized an enveloping procedure in
an effort to capture the effects of this soil variance in our assessments of foundation
stability and overall structure deformation.
Seattle’s bridge inventory is located within the Seattle Fault Zone, a relatively
high seismic zone. Peak ground accelerations for the upper level earthquake range from
0.43g to 0.47g. According to Geotechnical Reports, the recurrence interval for large
earthquakes with large ground surface deformation in the Seattle Fault Zone is on the
order of 3,000 to 5,000 years. This is much longer than the return periods of the upper
and lower ground level motions (100- and 1,000 years). As such, the risk posed by faultinduced
ground surface rupture for the ground motion hazard level has been categorized
as low to moderate. None of the bridges initially selected for investigation span currently
mapped splays, so near-fault effects were considered on a qualitative basis only.
Liquefaction
Liquefaction induced settlement has been identified with a high probability at all
of the bridge locations. However, since the vast majority of bridge foundations in these
areas are supported by piles, the risk of collapse due to this form of settlement is
considered low. The greater risk is posed by the deteriorated lateral support of the piles
during liquefaction.
The terrain at the majority of the selected bridge sites is relatively flat, for the
most part. In most cases the risk of lateral spread has been categorized as low, with the
exception of specific locations adjacent to the Fauntleroy Expressway.
Project Highlight - South Albro Place Bridge
The South Albro Place bridge spans over Airport Way South and BNSF and
UPRR railroad tracks and connects Stanley Avenue South with Swift Avenue South in
the area commonly referred to as Georgetown, just north of Boeing Field and west of I-5.
The bridge was originally constructed in 1931. (See Figure 2 and Photo 4).
Overall bridge length is approximately 172-meters, measured between abutments.
The bridge consists of three frame units separated by expansion joints and Bents 3 and 6.
Unit 1 superstructure, from Abutment 1 to Bent 3, is comprised of reinforced concrete
slab spans with drop-cap bent caps at Bents 1 and 2. The other two units’ superstructures
consist of reinforced concrete haunch girders with standard bridge deck. Each span has
three girders of varying depths, depending on adjacent span lengths.
All bents are supported on timber piles. Bents 1 and 2 are two-column bents, with
rectangular reinforced concrete column supported atop a truncated pyramid plinth that
extends down to the reinforced concrete pile cap. Bent 3 consists of three reinforced
concrete columns, with concrete counterforts. Bents 4 through 10 are all three column
bents. There are three reinforced concrete struts spanning between columns located at the
top of the plinth, mid-height, and at the capital. Bent 10 was similar to bents 4 through 9
except for the absence of the middle concrete strut.
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The tops of the columns of Bents 1 and 2 are integrally connected to the
superstructure. Each column of Bent 3 is connected to the bridge girders by means of a
mechanical pin-bearing assembly. At Bents 4 thru 10, the top of each column terminates
with a rocker bearing assembly embedded inside the trapezoidal concrete capital. Grout
was placed around the perimeter of the capital to nearly fill the space between the
reinforced concrete capital and the bridge girder.
The bridge appears to be in generally fair condition with minor, localized concrete
spalls and some cracking.
Displacement Demand Analysis
There were four distinct structural models associated with the as-built analysis of
South Albro Place Bridge. The “compression model” consisted of the entire bridge
modeled together assuming that the expansion joints at Bents 3 and 6 have closed.
Additionally, there were three independent frame “tension” models that were established
to determine the expected displacements in the event that an adjacent frame was not there
to provide restraint. The demands from the respective models were compared for each
bent and the largest value in each instance was used to determine the transverse and
longitudinal displacement demands.
FWHA guidelines (Ref. 4) were followed to provide initial soil springs for items
such as abutment backwalls, and pile footing springs, etc. The overall structural model
was then run with these initial springs. Abutment backwall springs were developed by
iterating until the combined longitudinal force at both abutments was less than or equal to
the expected soil passive resistance at only one end, since only one end of the bridge was
compressing soil at any given time. Additionally, the longitudinal displacement was
checked to assure that the gap between abutment backwall and superstructure was closed
and that the soil had been engaged. Transversely, the abutment forces were checked
against the footing’s sliding resistance.
Pile footing stiffnesses were iterated using WSDOT’S DFSAP program so that
the forces applied to the foundations and resulting displacements and rotations
determined from the SAP2000 analysis matched up with those used in the DFSAP model.
Seismic demands were determined from a multi-modal response spectrum
analysis using SAP2000 v.11. (See Figure 3) A sufficient number of modes were
included to account for at least 90-percent participation of the total mass. Modal
response contributions were combined using the CQC method. Response of the structure
was analyzed in two orthogonal horizontal directions and the results were combined
according to the SRSS rule. Vertical acceleration effects were not included in this
analysis.
The following assumptions were made for all response spectrum models:
1. Members were modeled with frame elements with 6 degrees of freedom at
each joint.
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2. The superstructure was modeled as a spine with 10 equal segments per
span. Superstructure frame elements were modeled at the composite neutral
axis.
3. End pier foundation stiffness was modeled with equivalent springs. Passive
pressure was included on the end pier backwall in accordance with §6.2.2.4
of the FHWA Manual (Ref. 4).
4. Interior piers were modeled as a frame with each column connected to a
rigid crossbeam. Column-to-crossbeam joint regions were modeled with
rigid links. A rigid link was also used to connect the crossbeam to the
superstructure. Piers were skewed from the superstructure spine to the
angles shown on the as-built drawings.
5. Columns were modeled with equivalent cracked stiffness. The reduced
stiffness values were determined using methods provided in the FHWA
Manual §7.3.2.1.
6. Interior pier foundation stiffness was modeled with equivalent 6 x 6
springs at each footing in accordance with WSDOT BDM (Ref. 9). Pile
foundations springs were generated using DFSAP. Spread footing springs
were generated by the Half-space Method, utilizing the formulas presented
in Section 6.2.2.1(b) of the FHWA Manual.
7. Mass was distributed in accordance with §7.3.1 of the FHWA Manual. The
inertia of live loads was not included.
8. A constant 5-percent damping coefficient was used for all modes.
Displacement Capacity Analysis
Albro Place Bridge was analyzed utilizing a Method D2 evaluation. The seismic
displacement capacity using non-linear static push-over analysis in SAP2000 v.11 was
determined from the following procedures:
• The un-factored dead load DC and DW were applied to the model as an initial
step. The resulting system displacement from the un-factored DC and DW
loading was the non-seismic displacement demand.
• Incremental lateral forces were applied to the system. A plastic hinge was
assumed to form at either the top or bottom of column and was formed when the
internal moments reach the idealized plastic moment capacity. Plastic hinging of
existing struts was permitted, however the struts were modeled with infinite
strength. The sequence of plastic hinging through the system was tracked until an
ultimate failure (collapse mechanism) was reached. The difference between the
system displacement at collapse and the non-seismic displacement demand was
the seismic displacement capacity.
• Separate pushover models were used to determine the longitudinal and transverse
seismic displacement capacity. The response spectrum bridge model was
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modified to include plastic hinges at the top and bottom of each bent column. The
model was then pushed in the longitudinal (parallel to centerline of structure)
direction to determine the longitudinal seismic displacement capacity. Each bent
was also modeled as a stand-alone frame model and pushed in the transverse
(normal to centerline of bridge) direction to determine the transverse seismic
displacement capacity.
• Mander's unconfined and confined concrete models (Ref. 6) were used to define
the concrete stress-strain curves. Expected concrete compressive strength was
equal to 1.50 times the specified 28-day concrete compressive strength (from the
as-built plans). Unconfined concrete compressive strain at maximum compressive
stress was taken as 2-microstrain. Ultimate unconfined concrete compression
(spalling) strain was limited to 5-microstrain.
• Park's model (Ref. 7) was used to define the reinforcing steel stress-strain curves.
Strain hardening of the main reinforcing bars was accounted for. Expected
reinforcing steel yield strength was equal to 1.10 times the yield strength (from
the as-built plans). Where reinforcing bar development lengths did not meet the
requirements of the FHWA Manual guidelines, the gross area of steel in the
section was reduced by a ratio equal to the development length provided, divided
by the development length required.
• Member strength capacities were generally determined from the AASHTO LRFD
code (Ref. 1), modified to exclude strength reduction factors and include ultimate
strength characteristics. Shear and moment capacities of reinforced concrete
sections in cross beams and foundations were calculated based on expected
member properties. Column sections were analyzed using the software XTRACT
which not only determines moment capacities, but additionally calculates
curvature capacities, which were used in the push-over analysis.
Individual items of the foundations, such as plinths, pile caps, and piles were analyzed as
member capacities described above. Overall foundation rocking capacities were
determined utilizing ultimate values of pile bearing capacity given in the Geotechnical
Report.
Section 7.8.2.7 of the FHWA Manual reads, “If the shear strength of the member
is less than the shear demand (based on flexural strength) the plastic rotation will be
limited. Two limiting cases are: (a) brittle shear, and (b) semi-ductile shear. These cases
are based on the shear strength relative to the flexural strength.” (Ref. 4) When the
initial shear strength of the member was less than the plastic shear demand, the member
was considered to be ‘shear-critical’ and would fail in a brittle manner with no plastic
rotation capacity. This type of failure was considered unacceptable unless the initial shear
capacity of the member was sufficient to resist the seismic loads elastically. Otherwise,
the member must be retrofitted to increase its shear capacity.
When the plastic shear demand lies between the initial shear capacity and the final
shear capacity of the member, the rotational capacity of the member was limited. If the
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limited rotational capacity of the member yielded a displacement C/D ratio that was less
than one, it may be possible to retrofit the member such that the final shear capacity of
the member exceeds the plastic shear demand and, thus, acts as a ductile member.
Provided the flexure-controlled rotational capacity was greater than the demand, the C/D
ratio of the retrofitted member would then exceed 1.0.
Table 1 below is a representative C/D Ratio table for South Albro Place Bridge in
the as-built condition, based on the Upper-Level Earthquake.
Table 1. Bent 7 (Bents 6, 8, 9, & 10 similar)
LOCATION / COMPONENT UNITS CAPACITY DEMAND
C/D
RATIO
Transverse Frame Displacement (mm) 25. 130. 0.2
Longitudinal Frame Displacement (mm) 185. 200. 0.9
Initial Column Shear (kN) 1100. 750. 1.5
Foundation Shear (Transverse) (kN) 1560. 450. 3.4
Foundation Shear (Longitudinal) (kN) 1560. 200. 7.8
Foundation Rocking (Transverse) (kN-m) 1600. 3700. 0.4
Foundation Rocking (Longitudinal) (kN-m) 2400. 2300. 1.1
Horizontal Struts (kN-m) 150. 890. 0.2
Expansion Joint Seat Width (Bent 6) (mm) 140. 775. 0.2
As illustrated in Table 1, the existing bridge exhibited several deficiencies under
the Upper-Level Event. Notably, the majority of the columns’ displacement capacities
were governed by semi-ductile shear. Additionally, the majority of the existing concrete
struts did not have the strength to resist the loads imparted to them through column
flexure. Furthermore, all of the existing foundations exhibited rocking behavior in the
transverse direction.
Recommended Seismic Retrofits
The following retrofits were recommended to mitigate the seismic vulnerabilities:
Concept 1 - Add steel column jackets with flared top and new steel top struts at
Bents 4 thru 10. Steel column jackets work to provide substantially greater displacement
ductility in the columns as well as addressing the columns’ shear capacity shortfall. No
longer limited by semi-ductile shear ductility capacity, the columns will be able to
accommodate the anticipated seismic deformations produced during the Upper Level
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Event. Additionally, the flared upper portion will not only provide confinement for the
rocker bearing assembly, but will also provide additional seat width, thereby reducing the
possibility of a girder becoming unseated from the column.
Concept 2- Remove existing counterforts at Bent 3 and column jacket similar to
Bents 4 thru 10. Removal of the counterforts allows the installation of steel column
jackets in a standard member for greater continuity and reliability. Also, all of the
benefits described in Concept 1 apply.
Concept 3- Add full depth reinforced concrete grade beams at Bents 3 thru 10.
Installation of full-depth reinforced concrete grade beams at these locations will provide
substantially improved performance of the foundation system when loaded in the
transverse direction. The foundations resistance to overturning will be greatly enhanced
by linking the individual foundations together, providing a much greater moment of
inertia for the pile group. Additionally, the unconfined concrete plinths will no longer be
subjected to bending moments, so plastic hinging of the jacketed columns will provide
the mechanism for deformation.
Longitudinal capacity of the pile group to resist overturning will not be as greatly
enhanced. However, utilizing additional passive resistance of the soil bearing on the
grade beam will provide additional overturning resistance.
Many of the concrete plinths extend 3.5-meters into the ground, so shoring will be
required for the construction of most of the grade beams. This is of special concern where
excavation takes place adjacent to railroad tracks, as both BNSF and UPRR have specific
requirements for structural shoring located next to tracks.
Concept 4- Add longitudinal restrainers at Bent 6 expansion joint. The in-span
rocker bearing at Bent 6 was particularly vulnerable. The location does not lend itself to
the installation of seat extensions, so we must rely on longitudinal restrainers to limit the
relative displacement of the adjacent frames such that the rocker bearing maintains
stability. We recommend placing these assemblies on the girders, which are quite
substantial and will provide excellent anchorage. The restrainers should be gapped to
allow for thermal expansion and contraction.
The restrainer assemblies are to be placed above existing railroad tracks, making
access to these locations both difficult and limited. The contractor will need to coordinate
this work with the representative railroad company to ensure safety was addressed.
Concept 5 – Add steel pipe transverse shear restrainers embedded in concrete
bolsters at Bent 6 expansion joint. We propose the installation of steel pipe shear
restrainers to provide shear resistance against differential transverse movements between
Units 2 and 3. Additionally, this measure would provide a modicum of additional vertical
support and may be viewed as seat extender.
Similar to the installation of the longitudinal restrainer units, this work would take
place in close proximity to active railroad tracks and requires careful coordination. It was
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possible that a work platform could be installed so long as railroad vertical clearances are
not jeopardized.
Concept 6 – Add seat extensions at Bent 3 to provide adequate seat length for
span 3. Originally, we intended to utilize longitudinal restrainers in this location, similar
to those described previously. However, installation of these units would be quite
difficult and require that the restrainer assemblies be rather tall and be mounted to the
underside of the bridge deck slab of span 3. Frame 1 does not rely on the stiffness of
frames 2 and 3 to keep the displacement demands at a level that the columns of Bents 1
and 2 can accommodate. As such, we recommend the installation of seat extensions onto
the ends of the girders to provide more than adequate seat width for span 3.
Concept 7 - Enlarge the pile caps and install additional piles at all bents to limit
rotations and potential differential settlement. The analysis indicates that foundation
rocking was likely to occur in the longitudinal direction for several bents. Additionally,
bents not retrofitted with full-depth concrete grade beams are likely to experience
foundation rocking for transverse loading as well. This was true for both the Upper and
Lower Level Events. Retrofitting the foundations with additional piles would greatly
enhance the foundation’s resistance to rocking loads and provide additional stability
below the ground surface. As a general rule, foundation retrofits are quite costly and
difficult to construct. Low overhead clearance limits the size and type of equipment that
can be utilized at the site. Also, the proximity of other features can severely impact the
installation of additional foundation elements. This was especially true for South Albro
Place Bridge. The proximity of existing railroad tracks at Bents 6 and 7 would require
that the affected rail lines be shored during construction.
Post-Retrofit Seismic Analysis
Table 2 below is a representative Post-Retrofit C/D Ratio table for South Albro
Place Bridge considering the demands of the Upper-Level Event.
Table 2. Bent 7 (Bents 3, 6, 8, 9, & 10 similar)
LOCATION / COMPONENT UNITS CAPACITY DEMAND
C/D
RATIO
Transverse Frame Displacement (mm) 1000. 70. 15
Longitudinal Frame Displacement (mm) 550. 115. 4.8
Foundation Shear (Transverse) (kN) 890. 600. 1.5
Foundation Shear (Longitudinal) (kN) 4400. 800. 5.5
Foundation Rocking (Transverse) (kN-m) 750. 310. 2.5
Foundation Rocking (Longitudinal) (kN-m) 6900. 8000. 0.9
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Conclusion
The tables shown above indicate several findings:
1. The addition of steel column jackets provided excellent displacement
ductility as expected.
2. The addition of the full-depth grade beams virtually eliminated the
likelihood of rocking in the transverse direction. Shear demands on the
foundations were increased, but only Bent 4 shows a C/D ratio less than one
for transverse foundation shear. Localized pile shear is not indicative of
structure collapse.
3. The addition of the full-depth grade beams slightly increased the
foundations’ resistance to rocking in the longitudinal direction. However,
without additional piles, all of the foundations are likely to experience
rocking in the longitudinal direction. Foundation rocking in the longitudinal
direction is not necessarily indicative of structure collapse.
In summary, the recommended retrofit concepts addressed the vast majority of
seismic vulnerabilities found in the as-built structure. The recommended optional retrofit
(Concept 7) would address the remaining item of longitudinal foundation rocking, but
due to both the significant expense as well as several constructability issues, this work
will be deferred.
Acknowledgements
Following are the City staff, the ERP Panel, the Senior Peer Review Consultants,
and our Subconsultants, without whom this program would not have been successful.
City of Seattle Engineering Department
• Bill Martin, Program Manager
• Kerensa Stoll, Deputy Program Manger
• John Buswell, Structural Engineering Manager
Expert Review Panel
• Paul Grant, PE, Pan GeoEngineers
• Roy Imbsen, D Eng, PE, SE, Imbsen
Consulting
• Michael Keever, PE, SE, Caltrans
• Frieder Seible, PhD, PE, University of
California – San Diego
• Charles Seim, PE, SE, Consulting Bridge
Engineer
• John Stanton, PhD, PE, University of
Washington
• Phillip Yen, PhD, PE, FHWA
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Senior Peer Review Consultants
• John Clark, PhD, PE, SE, Consulting Bridge
Engineer
• Lee Marsh, PhD, PE, SE, ABAM Engineers.
• Jim Schettler, PE, SE, Jacobs Civil.
Subconsultants
• Jacobs Civil
• Shannon & Wilson
• Civiltech
• Andersen Bjornstad Kane Jacobs, Inc.

References
1. AASHTO Subcommittee on Bridges and Structures (2007) LRFD Bridge Design
Specifications, Updated Seismic Provisions in LRFD Specifications, and 2008 Interims.
2. AASHTO Subcommittee on Bridges and Structures (2007) AASHTO Guide Specifications for
LRFD Seismic Bridge Design, Ballot Item 7, Wilmington, Delaware, (to be published ~
2008).
3. Applied Technology Council, ATC (1981) ATC-6, Seismic Design Guidelines for Highway
Bridges, Redwood City, CA.
4. Buckle, I.G. et al (2006) Seismic Retrofitting Manual for Highway Structures, Part I - Bridges,
MCEER-06-SP10 and FHWA-HRT-06-032.
5. Federal Highway Administration, FHWA (1995) Seismic Retrofit Manual for Highway
Bridges, Publication No. FHWA-RD-94-052.
6. Mander, Priestly, Park, “Theoretical Stress-Strain Model for Confined Concrete,” Journal of
Structural Division, ASCE, Aug, 1988.
7. Park, Paulay, Reinforced Concrete Structures, Wiley, 1975.
8. Priestly, Seible, Calvi, Seismic Design and Retrofit of Bridges, Wiley, 1996.
9. WSDOT, Bridge Design Manual LRFD, May 2008.
FIGURES & PHOTOS
Figure 1. Project Vicinity Map Figure 2. South Albro Place
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Spectral Acceleration, Sa (g)
1.2
1.0
0.8
0.6
0.4
0.2
0.0
0
0.5
Response Spectra - Albro Place Bridge
T=0.71 - 0.74s
1
1.5
Period, T (sec)
Figure 3. Response Spectrum Curves for Albro
Photo 4. South Albro Place Bridge
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2
Site Class C
1000-YR / 100-YR
PGA = 0.46g / 0.15g
S S = 1.02g / 0.33g
S 1 = 0.34g / 0.10g
Fa = 1.0 / 1.2
Fv = 1.46 / 1.70
S DS = 1.02g / 0.40g
S D1 = 0.50g / 0.17g
T 0 = 0.10s / 0.09s
T S = 0.49s / 0.43s
2.5
Upper Level Event
Lower Level Event
3