Louis Berger Group Inc. Kajakai Hydroelectric Project ... - Afghan

Louis Berger Group Inc.
Afghanistan
Kajakai Hydroelectric Project
Condition Assessment
Dam Safety Assessment Report
April 2004
P15300
Acres International Corporation
Amherst, New York

Table of Contents
List of Tables
List of Figures
Executive Summary
1 INTRODUCTION ............................................................................................................. 1-1
1.1 BACKGROUND............................................................................................................. 1-1
1.2 DAM CONDITION ASSESSMENT OBJECTIVES...............................................................1-1
1.3 KAJAKAI DAM ASSESSMENT.......................................................................................1-2
2 THE KAJAKAI DAM.......................................................................................................2-1
2.1 DESCRIPTION OF EXISTING STRUCTURES ....................................................................2-1
2.2 HISTORY .....................................................................................................................2-1
2.2.1 Original Project Implementation – 1946 to 1956..................................................2-1
2.2.2 Powerhouse Construction - 1970 (approximately) to 1975...................................2-5
2.2.3 Final Dam and Spillway Works – 1978 to 1980 (approximately) .........................2-6
2.3 DESIGN .......................................................................................................................2-6
2.3.1 Reservoir and Spillway Gates Elevation Criteria..................................................2-7
2.3.2 Embankment Dam .................................................................................................2-8
2.3.3 Dam Foundation....................................................................................................2-3
2.3.4 Spillway ................................................................................................................. 2-4
2.3.5 Emergency Spillway ..............................................................................................2-6
2.3.6 Tunnels ..................................................................................................................2-6
2.3.7 Powerhouse ........................................................................................................... 2-6
3 DATA COLLECTION......................................................................................................3-1
3.1 REPORTS AND PAPERS.................................................................................................3-1
3.2 DRAWINGS..................................................................................................................3-2
3.2.1 Morrison Knudsen Afghanistan Inc. Drawings .....................................................3-2
3.2.2 International Engineering Company, Inc. Drawings ............................................ 3-2
3.2.3 Hitachi Shipbuilding and Engineering Co. Ltd. ....................................................3-2
3.2.4 Water and Power Development Consultancy Services (India) Ltd........................3-3
4 SITE INSPECTION AND CONDITION ASSESSMENTS ........................................... 4-1
4.1 INTRODUCTION ........................................................................................................... 4-1
4.2 GEOLOGY....................................................................................................................4-1
4.2.1 Kajakai Project Area Description .........................................................................4-1
4.2.2 Regional Geology ..................................................................................................4-1
4.2.3 Engineering Geology of the Damsite and Spillway Area ......................................4-2
4.3 DAM............................................................................................................................4-5
4.3.1 Crest ......................................................................................................................4-6
4.3.2 Upstream Face ......................................................................................................4-7
4.3.3 Downstream Face..................................................................................................4-8
4.3.4 Seepage..................................................................................................................4-9
4.3.5 Embankment Instrumentation.............................................................................. 4-10
4.4 SERVICE SPILLWAY................................................................................................... 4-10
4.4.1 Concrete Structure............................................................................................... 4-10
4.4.2 Geotechnical Aspects of the Spillway.................................................................. 4-11

Table of Contents - 2
4.5 EMERGENCY SPILLWAY ............................................................................................ 4-13
4.6 WATER CONVEYANCE STRUCTURES......................................................................... 4-13
4.7 RESERVOIR ............................................................................................................... 4-14
5 ASSESSMENT OF DAM..................................................................................................5-1
5.1 EMBANKMENT DAM STABILITY..................................................................................5-1
5.1.1 Stability Analyses...................................................................................................5-1
5.1.2 Geotechnical Parameters ......................................................................................5-1
5.1.3 Reservoir Levels ....................................................................................................5-2
5.1.4 Seismic Parameters ...............................................................................................5-2
5.1.5 Analysis Results.....................................................................................................5-4
5.2 SEEPAGE AND DRAINAGE CONTROL ........................................................................... 5-4
5.3 DEFORMATION AND CRACKING ..................................................................................5-5
5.4 SURFACE EROSION......................................................................................................5-6
5.5 LIQUEFACTION POTENTIAL .........................................................................................5-6
5.6 INSTRUMENTATION .....................................................................................................5-7
6 SPILLWAY ASSESSMENT.............................................................................................6-1
6.1 CONCRETE STRUCTURE...............................................................................................6-1
6.2 SLOPE STABILITY........................................................................................................6-1
6.3 SEEPAGE CONTROL.....................................................................................................6-1
6.4 FOUNDATION SHEAR STRENGTH................................................................................. 6-2
6.5 OUTLET CHANNEL EXCAVATION ................................................................................ 6-2
6.6 EMERGENCY SPILLWAY ..............................................................................................6-2
7 SUMMARY AND RECOMMENDATIONS...................................................................7-1
7.1 SUMMARY OF FINDINGS..............................................................................................7-1
7.2 RECOMMENDED ADDITIONAL INVESTIGATIONS AND STUDIES....................................7-2
7.3 RECOMMENDED REMEDIAL WORK ............................................................................. 7-7
References
Appendix A Photographs
Appendix B Seismicity Review
Appendix C Dam Stability Analyses – Analysis Criteria

List of Tables
No. Title
2.1 Key Elevations for the Kajakai Dam
2.2 Reservoir Data for 1956 to 1976
2.3 Reservoir Volumes
2.4 Fill Materials Properties
2.5 Spillway Design Elevations
4.1 Description of Major Discontinuity Sets
in the Kajakai Dam and Spillway
5.1 Representative Materials Properties Assumed
for Stability Analyses of the Kajakai Dam
5.2 Reservoir and Tailwater Level Used for
Stability Analyses of the Kajakai Dam
5.3 Peak Ground Acceleration and Sustained Seismic Coefficient Values
5.4 Kajakai Dam Stability Analysis Results

List of Figures
No. Title
1 Location Map
2 Kajakai Dam and Reservoir
3 General Layout – Kajakai Dam and Spillway
4 Kajakai Dam – Main Section
5 Kajakai Dam and Spillway Discontinuity Survey
B1 Seismic Hazard of Afghanistan
B2 Earthquake Epicentre Locations

Executive Summary

Executive Summary
i
The Kajakai Dam is a 90-m high embankment dam with an uncontrolled open channel
spillway, which was constructed on the Helmand River in Afghanistan during the
early 1950s to provide river control and irrigation benefits. A 33 MW powerhouse
was added to the project in 1975. Work on the planned spillway gates, emergency
spillway and raising the dam crest commenced during the late 1970s but construction
activities ceased during the Russian occupation and these facilities were never
completed. Consequently, the reservoir has never been impounded to its design level
of el 1045 m. Maximum reservoir levels reached to only approximately el 1037 m
during the 50 year operation of the project.
The Kajakai Dam was originally part of the irrigation systems controlled by the
Helmand and Arghandab Valley Authority (HAVA). Administrative control of the
facility currently rests with the Ministry of Irrigation, Water Resources and
Environment (MIWRE). MIWRE now plans to upgrade the Kajakai structures to
increase the reservoir capacity, irrigation capacity and power generation facilities.
Given the age of the dam and the largely undocumented operational history, a dam
condition study has been commissioned as first step in the planned remedial and
improvement works.
The dam condition study assessments concluded the following:
• Project documentation is incomplete. In general, many details of the original dam
and powerhouse design are available. No information about the final design of the
spillway gate structure, emergency spillway or planned raising of the dam crest
was available for review. No “as-built” information of the partially constructed
civil structures of the aborted 1979-80 construction program is available.
• The dam is in good condition. It is possible that there has been some partial
desiccation of the upper levels of the core above the past impoundment levels, but
this is not a significant impediment to raising the reservoir level. Available
information indicates that the reservoir can be safely filled to the design
impoundment level of el 1045 m, provided that the dam is effectively monitored
and the initial filling is carried out a carefully controlled rate.
• Slope stability analyses of the dam embankment indicate that it is acceptably
stable and will withstand the effects of full impoundment of the reservoir for the
normal and rapid drawdown loading cases. The pseudostatic seismic stability
analyses indicate that the upstream slope has marginal stability for Maximum
Design Earthquake seismic loadings. This consideration, coupled with the
possibility of liquefaction of the foundation soils, necessitates future earthquake
stability review of the dam.

ii
• The existing erosion protection on the upstream face of the dam has deteriorated
over the past 50 years. Some slope dressing work is required to restore the riprap
slope protection. A few areas in the downstream face also require some minor
slope dressing work.
• Only limited seepage could be observed in the right bank of the river and in the
left bank tunnels in the area downstream from the dam. This is evidence of the
effectiveness of the dam and its grout curtain. There is, however, an unexplained
hole in the el 980 m bench roughly on line with an area where there was reported
seepage from the toe in the 1950s. This feature should be assessed in the field and
investigated further.
• It is not clear if the crest of the dam was raised during the 1979-80 construction
program. The current crest elevation and the profile of the top of the core are
uncertain.
• There are no operable means or instrumentation to measure settlement and/or
displacement within the dam or hydraulic gradients within the dam core.
• The uncompleted spillway concrete structure is in fair to good condition and can
most probably be incorporated into the final structure.
• No records are currently available for the design of the spillway structure which
was started in the late 1970s. Design construction details must be found and
testing of the structural fabric (concrete and rebar) must be carried out before a
meaningful technical review of the spillway structure can be completed.
• The excavated spillway channel rock slopes have suffered some deterioration over
the years. These are in a hazardous condition in some areas above future work
sites for the planned gate installations. Some remedial works, including scaling of
loose blocks and localized rock support are needed.
Additional studies are recommended to enable a more complete assessment of the
dam and related structures. The recommended studies include the following:
• Topographical surveys of the dam crest and spillway areas.
• Locate copies of missing design drawings and as-built details from the original
engineering consultants and Afghanistan government sources.

iii
• Carry out further assessment of installed instrumentation and the installation of
deformation monitoring bench marks. New piezometers may be needed , pending
the results of the review of the existing system.
• Test pit investigations in the crest of the dam and in a possible sinkhole on the el
980 downstream berm.
• Seismicity study and further stability analyses on the dam to determine the
potential for foundation liquefaction and the magnitude of earthquake induced
deformations.
• Hydraulic and hydrologic engineering reviews of the spillway designs.
• Review of the reservoir rim.

1 Introduction

1 Introduction
1.1 Background
The Kajakai Dam and its associated structures, were constructed on the Helmand
River during 1951 to 1953. A powerhouse was commissioned in1975. Additional
construction works were started during 1978 to 1980 to install spillway gates and
raise the reservoir to the original design level. This work was terminated in 1980 after
the Russian invasion and none of the new facilities were ever finished. The existing
facility consists of a 90-m high embankment dam, an uncontrolled open spillway, a
partially excavated emergency spillway, irrigation discharge facilities and a
powerhouse with a rated capacity of 33 MW.
The Kajakai Dam was originally part of the irrigation systems controlled by the
HAVA. Administrative control of the facility currently rests with the MIWRE.
MIWRE now plans to upgrade the structure to increase the reservoir capacity,
irrigation capacity and power generation facilities. Given the age of the dam and the
largely undocumented operational history, a dam condition study has been
commissioned as first step in the planned remedial and improvement works. The
current reservoir has never been impounded to a level higher than about el 1037 m.
The new spillway gates will result in higher reservoir levels of 1045 m. Acres
International (Acres) was retained by Louis Berger Group Inc. (LBG) to undertake a
condition assessment of the existing powerhouse and the dam
This report presents the results of the civil and geotechnical assessment of the Kajakai
Dam. Hydrologic and hydraulic assessments of the spillway capacities have not been
included in this assessment.
1.2 Dam Condition Assessment Objectives
The objectives of the condition assessment review are as follows:
• assessment of the condition of the dam and its components
• performance of a detailed site inspection
• identification of more detailed investigations required, if necessary
• identification of necessary repairs and/or continuing maintenance needs
• assessment of the impact on the dam of raising the reservoir to the original design
levels after the spillway gates are installed.
Specifically, the condition assessment of the dam comprises a procedural evaluation
of its ability to withstand the forces that could be expected to act on it during the
lifetime of the structure. The Kajakai Dam, which is more than 50 years old, has
Comment: Not sure this is correct –
Ithought it was with the Ministry of
Water and Power – did Allan give you
this info?

1-2
never impounded the reservoir to its full design level. Particular attention is therefore
paid to judging its ability to operate with a full reservoir when the gate structures are
completed as planned. The age of the structure and the unfinished spillway facilities
are important components in this review. The bulk of the evaluation considered the
geotechnical aspects of the dam and spillway. The engineering procedures used for
this evaluation are compatible with those specified by the Federal Energy Regulatory
Commission (FERC), “Engineering Guidelines for the Evaluation of Hydropower
Sites”.
The lack of construction, design and operational information present a handicap in
giving a full condition assessment of the dam at this time.
1.3 Kajakai Dam Assessment
The Kajakai dam is located on the Helmand River in the Afghanistan province of
Helmand. It impounds a 27-km long reservoir which provides flood control,
irrigation water and hydroelectric power for the region.
This report comprises the following sections:
• Executive Summary.
• Section 1, Introduction – introduction and explanation of approach.
• Section 2, Kajakai Dam – information on the Kajakai Dam and related structures.
• Section 3, Data Collection – details of the initial data review including the type of
documents reviewed.
• Section 4, Site Inspection and Condition Assessments – details of the site
inspections carried out.
• Section 5, Dam Assessment – assessments of the dam.
• Section 6, Spillway Assessment – assessments of the spillways.
• Section 7, Summary and Recommendations – provides a summary of
recommended investigations and some remedial measures.

2 The Kajakai Dam

2 The Kajakai Dam
2.1 Description of Existing Structures
The existing facilities consist of an embankment dam, a surface powerhouse,
irrigation discharge facilities, an ungated open channel spillway and a partially
completed emergency spillway. The dam retains a reservoir of 2.35 km 3 . Gated
intakes for the powerhouse and irrigation discharge structures are located near the left
abutment of the dam. The project location and a plan of the reservoir are shown on
Figures 1 and 2 respectively. The project layout is shown on Figure 3. Design
details of the major structures are given in the following paragraphs. Photographs of
the various structures are presented in Appendix A.
The major components of the scheme and the approximate construction end dates are
as follows:
• embankment dam, 1953
• irrigation intake and outlet works, 1953
• ungated spillway with overflow weir, 1953
• powerhouse, power intake and penstock, 1975
• uncompleted spillway concrete gate structure, 1979-80
• uncompleted emergency spillway, 1979-80
• fabrication and delivery of spillway gate components (currently unassembled),
1979-80.
2.2 History
The Kajakai Dam was initially part of a larger irrigation and river control scheme
which also encompassed the Boghra Canal and Arghandab Dam projects. Work on
these schemes commenced in 1946. Three stages of detailed design and construction
were carried out for the Kajakai Dam.
2.2.1 Original Project Implementation – 1946 to 1956
Field studies and surveys for the Kajakai Project commenced in 1946.
Preliminary designs were carried out by 1950. Investigations and design work
were carried out by Morrison-Knudson Afghanistan and its subcontractor design
firm, International Engineering Company Inc. Helmand Construction Company
(HCC) was the contractor. Construction commenced after 1950 and the bulk of
the structures were built during the period 1951 to 1953. Final design reports
were issued in 1956.

Kajakai Dam
Figure 1
Kajakai Dam Condition Assessment
Location Map

0 1 2 3 4
Kilometres
Helmand River
Kajakai Dam
Service Spillway
Downstream Reservoir
Rocky Narrows
Upstream Reservoir
Alluvial Fan
Figure 2
Kajakai Dam Condition Assessment
Kajakai Dam and Reservoir

0
Metres
100 200
Source:
International Engineering Company Inc.
“Final Design Report on Kajakai Dam, Arghandab Dam and Boghra Canal Projects”
PARTIALLY EXCAVATED
EMERGENCY SPILLWAY
(APPROXIMATE LOCATION)
Figure 3
Kajakai Dam Condition Assessment
General Layout - Kajakai Dam and Spillway

2-5
The following items were constructed during the original construction program in
the early 1950s.
• embankment dam with an original design crest level at el 1050 m.
• ungated open channel spillway with 100 m long weir in the spillway (crest
level at el 1033.5 m)
• two unlined diversion tunnels
• irrigation intake structure, located upstream of the left abutment of the dam.
• three steel conduits for irrigation water supply in the outer or left diversion
tunnel, which extend downstream from the concrete plug structure.
• irrigation discharge structure, which included three control valves.
First impoundment of the reservoir occurred in 1953. Following spring runoff in
June 1953, the reservoir filled to el 1031.5 m or 2.0 m below the spillway crest.
The first spillway discharge occurred on April 18, 1954. The maximum discharge
of that year occurred on May 3 when the reservoir rose to el 1035.8 m with a
corresponding discharge of 678 m 3 /sec. Available records, after 1954, show that
spillway discharge occurs for about one to two months month during the spring
and early summer of most years.
2.2.2 Powerhouse Construction - 1970 (approximately)
to 1975
International Engineering Company Inc. and Harza Engineering carried out
further design work for the powerhouse, spillway and dam. Fishbach–Oman was
retained to construct the powerhouse and this structure was completed and
commissioned in 1975. The facilities constructed and/or installed during this
second construction program include the following.
• power intake structure and inclined power tunnel leading to the inner or right
diversion tunnel
• installation of a new concrete plug in the inner or right diversion tunnel,
downstream from the inclined power tunnel
• 3.6 m diameter steel penstock located within the inner or right power tunnel,
extending from the new concrete plug to the powerhouse,
• three bay surface powerhouse
• installation of turbines and generators in the outer two bays (Units 1 and 3)
• foundation preparation for a powerhouse extension.
A series of conceptual drawings were prepared for the spillway gate structure
during this period.

2-6
2.2.3 Final Dam and Spillway Works – 1978 to
1980 (approximately)
International Engineering Company Inc. carried out further design work on the
spillway structures at some time after 1975. This work included final plans for the
spillway gate structure, a fuse plug emergency spillway and raising of the dam
crest level. Details of this work are not available at the present time.
Construction work was carried out between approximately 1978 and 1980. It is
known that the following work items were performed before the project was
terminated at the time of the Russian occupation in 1980.
• excavation of the spillway to final grade as required
• excavation of a gallery, possibly for foundation drainage, into the lower right
abutment of the gate structure
• partially construction of the foundation for the ogee structure of the concrete
gate structure
• carried out partial excavation and dental concrete backfill work in a faulted
karstic feature on the upstream side in the middle of the spillway structure
• partially excavated the emergency spillway
• fabrication of the steel gate components by Hitachi Shipbuilding and
Engineering Co. Ltd.
• delivery of the steel gate components to the Kajakai site.
The steel gate components are still present on the site and are in good condition.
No as-built details are available for the civil works. It appears that the top of the
concrete spillway ogee structure is in the range of approximately el 1035.0 m to
1035.5 m. This compares with the design ogee crest level of el 1037.5 m. No
design data of the fuse plug or emergency spillway are available for review.
It is understood that the Harza design criteria specified that the dam crest should
be raised by about 2 meters as part of the spillway gates installation procedure.
To date, no information has been found on this issue. It is possible that the dam
was raised during either the 1975 powerhouse construction work or the 1978 to
1980 spillway gates construction program. This should be confirmed by a
topographic survey and further detailed review of old construction records.
2.3 Design
Available design reports note the following design and construction related items,
which must be considered during the condition assessment work.

2-7
2.3.1 Reservoir and Spillway Gates Elevation Criteria
Reservoir levels summaries are taken from the “The Kajakai-Kandahar Power
Supply System”, Ministry of Water and Power, August 2003. A summary of dam
crest, reservoir and tailwater levels are given in Table 2.1.
Table 2.1
Key Elevations for the Kajakai Dam
Description Elevation (m)
Dam Crest (original)
1050.0
Normal Present (with uncompleted Approx 1035 to
Maximum gate structure)
1035.5 (estimated)
Reservoir Planned (with completed
1045
Levels gate structure)
Original in 1953 1033.5
Minimum operating reservoir level (design) 1012.0
Actual minimum operating level 1008.0
Tail water level 978.0
As can be seen, completion of the spillway gates as planned will increase the
reservoir level by about 10 m. The elevation of the current dam crest is subject to
confirmation by field survey.
The operation history of the Kajakai spillway and reservoir for 1956 to 1976 is
summarized in Table 2.2.
Table 2.2
Reservoir Data for 1956 to 1976
Reservoir Elevation Max. 24 Hour Flow Flow Over Spillway
Year Min. Max into Reservoir m 3 x 10 6 From To
1956 1015.13 1036.66 110.5 April 5 August 5
1957 1011.70 1037.50 146.7 April 5 July 18
1958 1004.00 1035.29 61.2 April 14 June 14
1959 1002.79 1035.25 54.1 April 15 June 12
1960 999.72 1035.54 78.8 May 7 June 18
1961 996.77 1035.00 85,6 April 27 June 19
1962 1000.48 1027.17 43.3 No Spill No Spill
1963 1005.62 1034.04 63.6 June 4 June 26
1964 1002.01 1036.52 92.1 April 22 June 12
1965 1010.53 1036.53 99.3 April 12 July 11
1966 1005.15 1025.74 45.0 No Spill No Spill
1967 1005.47 1037.35 203.2 April 19 July 1
1968 1012.42 1035.79 85.3 April 25 June 25
1969 1017.69 1036.78 164.5 April 3 June 28
1970 1006.36 1021.88 39.2 No Spill No Spill
1971 994.93 1018.94 26.0 No Spill No Spill

2-8
Reservoir Elevation Max. 24 Hour Flow Flow Over Spillway
Year Min. Max into Reservoir m 3 x 10 6 From To
1972 989.51 1035.81 68.2 April 10 June 20
1973 978.00 1035.29 67.6 April 24 June 17
1974 977.04 1034.06 30.2 May 6 May 31
1975 1006.82 1035.93 98.0 April 17 June 28
1976 1014.95 1037.46 156.2 April 14 July 3
Reservoir volumes for ungated and gated spillway cases are summarized in
Table 2.3.
Table 2.3
Reservoir Volumes
Reservoir Volume (m 3 )
Present
Case
(No Gates) With New Gates
Total Storage Volume 1,715,774,752 2,725,997,270
“Active” Volume
(with minimum pool at el 1012 m)
1,134,804,293 2,145,026,811
Increase in Active Volume n.a. 1,010,222,518
2.3.2 Embankment Dam
The dam is an earth and rockfill structure with an original design height of 90 m
above the streambed. The original crest was 10 m wide, 270 m long and had a
design level at el 1050 m. It was planned to increase the crest level by a few
metres during the 1978 to 1980 construction program. It is not clear if this dam
heightening was ever carried out and the current dam crest elevation is uncertain.
The overall dam slopes are 2.5:1.0 and 3.0:1.0 upstream and 2.0:1.0 downstream.
The dam has a compacted central impervious core flanked by compacted semipervious
and pervious zones. Riprap slope protection consists of dumped rockfill
on the upstream and downstream faces of the dam. The impervious core is
founded upon bedrock in a cutoff trench which was excavated through the
riverbed alluvium. The upstream and downstream semi-pervious and pervious
shells are founded upon the natural riverbed alluvium in the old river bed area.
The details of the dam zoning and design are shown on Figure 4. This figure is
adapted from the Morrison Knudson 1956 design report. Very little information is
available on the construction materials used in the interior of the dam. Table 2.4
shows average material gradations as determined by laboratory testing data found
in the Kajakai powerhouse. This testing was carried out in 1976 on samples
obtained from the borrow areas.

0
Metres
25 50
Source:
International Engineering Company Inc.
“Final Design Report on Kajakai Dam, Arghandab Dam and Boghra Canal Projects”, 1956
Figure 4
Kajakai Dam Condition Assessment
Kajakai Dam - Main Section

Table 2.4
Fill Materials Properties
2-2
Average per cent passing
Material No. 200 mesh No 8 mesh ¼ in. mesh 1 ½ in mesh
Impervious 67.3 82.4 88.7 98.8
Semi-pervious 33.6 50.0 56.3 90.0
Significant design and construction details are summarized as follows.
• The overall slope of the upstream face is 2.5:1 for the upstream face and 2.0:1
for the downstream face. Each face of the dam is made up of a series of 6.5 m
wide berms, which have a vertical spacing of 10.0 m. The slopes between the
berms consist of rock fill which was dumped to its angle of repose, subject to
some reshaping at a later date, in most cases.
• The original crest grade included an upward camber along the centreline to
accommodate post construction embankment settlement. The maximum
camber is 1.5 m at the maximum section, Sta. 0+280, where the crest elevation
was constructed at el. 1051.51 m. This maximum camber was graded
downwards to zero at each of abutments (Sta. 0+174 and Sta. 0+447 m).
• The outer shells are constructed of dumped and sluiced, uncompacted rockfill.
The remainder of the embankment dam, including the core, semi-pervious and
transition zones, was placed in compacted lifts.
• The downstream rockfill shell incorporates the downstream cofferdam. This
cofferdam consisted of dumped rockfill which was placed at its angle of
repose. The three lowest berms of the downstream face of the existing dam,
the el 990, el 980 and 970 berms, are part of the original cofferdam
embankment.
• The rock fill shells were constructed of blasted limestone rock from the
spillway and tunnels. Not all of blasted rock from the spillway could be used
in the dam rock fill shells as planned. A large proportion of the blasted rock
was reportedly contaminated with clayey fault material and was considered
unsuitable for use as rockfill. A special quarry was also used as a
supplementary source of rock fill. The location of this quarry is not known.
• The transition zones and filters were constructed of unprocessed granular
material, which was obtained from borrow areas upstream and downstream of
the dam.

2-3
• A number of borrow areas, which were mostly located in the present reservoir
area, provided impervious fill for the core of the dam. A considerable effort
was made to ensure adequate moisture control when borrowing this material.
Borrow pits were reportedly terraced and irrigated well in advance of final
excavation in order to achieve the desired moisture contents in the fill.
• A total of 13 plastic pneumatic piezometer tips were embedded in the
embankment at Station 0 + 290. The majority were located in the impervious
core. Two plastic tubes from each tip were carried through the embankment to
individual compound-element Bourdon-type gauges, which were located in a
terminal cabinet on the left abutment at the toe of the dam. No details of the
piezometer installations or monitoring results are available.
2.3.3 Dam Foundation
The dam is founded on a slightly karstic dolomitic limestone rock mass. The
foundation contains systematic sets of natural fractures and a number of faults.
Many of the various discontinuities have been affected by karstic solutioning.
Clay infillings are common in many fractures and cavities.
The designers of the dam implemented a number of works to treat potential
leakage under the dam and to mitigate other foundation problems. The following
information is summarized from the Morrison Knudson final design report (1956)
and from a number of monthly progress reports which were found at the project
site. This data is relevant in considering the current dam condition and long term
safety.
• The grout curtain was constructed 20 m deep into the rock along the valley
floor and in the lower abutments. It is 12 to 16 m deep in the upper
abutments. The grout curtain is single line and was constructed with 10 m
spaced primary holes. Secondary, tertiary and, where necessary, quaternary
holes were installed by split spacing methods until grout closure was achieved.
• Progress reports indicate that the grouting engineers made special efforts to
wash out clay seams encountered by the grout holes at some locations. Details
of this work are not available.
• The grout curtain under the dam was extended to tie into the radial curtains
around the tunnels.
• Excavation and rock treatment was carried out to treat a 15 m wide buried
gorge in the bedrock, which was found beneath the streambed. The bottom of

2-4
the gorge was found to be infilled with dense granular alluvium. When the
alluvium was removed, the old channel bottom was seen to contain rock
pinnacles, potholes and sinuous troughs up to 1.8 m in depth. After cleaning
out the unconsolidated material, the rock surface was covered with gunite and
depressions were filled with lean concrete in order to bring the surface up to a
general level which would permit the placement of impervious fill.
• A five metre wide fault zone, aligned parallel to the river, was encountered
below the riverbed. This feature is aligned parallel to the buried valley
described above and is located “slightly over to one side” of it. Extensive
treatment works were carried out for this feature. These included a 30 m deep,
5 m wide by 6 m long shaft. The shaft was backfilled with 1,343 m 3 of
concrete. Six grout holes were extended 15 m beyond the bottom of the shaft.
In addition to the shaft, the entire length of the fault zone under the impervious
core was cleaned to a depth of 1.2 to 1.5 m and backfilled with concrete.
• A second fault, about 1.8 m wide was located on the abutment slope. This
fault reportedly crosses the dam axis at an angle of about 28 to 30 degrees.
Two shafts were excavated into the fault and backfilled with concrete. The
first shaft was 21 m deep and was located about 16 m downstream from the
dam axis. The second shaft was located along the grout curtain and was 8.5
deep.
2.3.4 Spillway
The spillway consists of a 101 m wide, unlined open cut channel, which is located
about 0.8 km from the right abutment of the dam. The spillway floor is cut
entirely in bedrock to approximately el 1032m. Presently, the spillway contains a
partially completed concrete ogee structure, which was intended to be the base of
an eight bay gated spillway structure. The specified ogee crest level was el 1037
m, but an examination of the uncompleted structure suggests that its present “asbuilt”
crest is somewhere in the range of el 1035 m to 1035.5 m.
The designers of the dam planned to construct the spillway in two stages. During
initial construction in the early 1950s, a weir, with a crest elevation of 1033.5 was
constructed in the open cut excavation. It was planned to operate the spillway in
this condition for the initial years of the dam operation. A gated spillway was
planned from the outset and work on this installation commenced in the late
1970s. This structure was never completed.
The spillway is excavated in sound dolomitic limestone. The 1971 International
Engineering Company Inc. 1971 feasibility report indicated that early geological
investigations delineated a fault zone running along the centre of the spillway,

2-5
oriented approximately parallel to the spillway axis. The fault zone was
determined to be about 2 m wide and would require special treatment in the
foundation of the concrete structure.
Initial layouts specified an eight bay gated spillway. A 12.19 m wide, 11.22 m
high radial gate would be installed in each spillway bay. Spillway elevation data
are presented in Table 2.5.
Table 2.5
Spillway Design Elevations
Description Elevation (m)
Bridge deck 1050.0
Weir crest (new weir) 1037.0
Old channel invert 1032.0
Approach channel invert 1029.6
The International Engineering Company Inc. 1971 feasibility design report
entitled “Feasibility Study for Installation of Spillway Gates at Kajakai Reservoir,
Afghanistan”, indicated the following geotechnical design considerations for the
new gates structure:
• Locate gates far enough upstream to minimize erosion of the gravel deposit in
the natural channel downstream of the spillway excavation. There are
economic trade-offs between erosion considerations and cost. It is cheaper to
keep the gates in a downstream location.
• Need a grout curtain under the floor and in the abutment walls.
• The structure requires a downstream apron and blasted shear key under the
gate foundation.
• Rock anchors are required in the abutment areas and through the floor slab.
• Treatment is required for the fault in the foundation. The design report
indicated the need for 20 to 30 m of excavation and concrete replacement.
• Additional excavations needed at exit of the natural discharge channel into
main river in order to protect the powerhouse and tailrace area. This work had
been planned as part of the original design, but was never carried out.

2.3.5 Emergency Spillway
2-6
A partially excavated emergency spillway is located between the main spillway
and the right abutment of the dam. It appears that work was suddenly halted on
this excavation as piles of blasted rock muck, uncharged blast holes and derelict
excavation equipment are still in place. As it presently exists, this excavation is
not deep enough to provide any spillage capability for the dam.
It is understood that a fuse plug embankment dam was planned to be constructed
in this spillway. No drawings or other design details for the emergency spillway
were available at the time of the preparation of this report.
2.3.6 Tunnels
Two tunnels were constructed to provide river diversion during dam construction.
These tunnels are unlined 10 m wide horseshoe shaped excavations, which are
spaced about 60 m apart. After completion of the dam, these tunnels were used as
water conveyance structures for the irrigation works and power flow.
The outer or left tunnel is used for the irrigation works. The irrigation intake is
located in the reservoir and includes a submerged trashrack with stop log slots for
emergency unwatering. A concrete plug was installed roughly on line with the
dam axis. Three steel conduits extend from this plug to the irrigation regulating
valves, which are located several metres downstream from the outlet portal of the
tunnel.
The inner or right tunnel contains the power penstock. The tunnel contains a
concrete plug, which is located approximately on line with the dam axis. The
3.6 m diameter steel penstock extends from the tunnel plug to the powerhouse.
Provision has been made to install a second penstock for a future powerhouse
extension.
2.3.7 Powerhouse
The powerhouse is a surface structure, which contains two vertical shaft Francis
turbines and generators. These units have a combined rated output of 33 MW.
Provision has been made for the future installation of a third unit in the existing
powerhouse.
It is understood that some foundation preparation has been made for an upstream
extension of the powerhouse.

3 Data Collection

3 Data Collection
An important step in the assessment process was to carry out a detailed review of
available information. As part of this process the following documents from several
sources were examined:
• site layout drawings
• design and construction drawings, where available
• old correspondence files at the site
• topographic maps
• early water level and stage-discharge data which was found at the site
• design report
• published technical papers
• construction contract documents.
Unfortunately, the project data is incomplete and is not located in a central repository.
Records have been obtained from a number of sources, including the Kajakai site,
Morrison Knudson and the Louis Berger Group. There is a large collection of
construction drawings and field correspondence in the Kajakai powerhouse in lower
level of the irrigation discharge structure. This site data is unorganized and locating
relevant material was fortuitous. A number of important documents and drawings are
still not available. Copies were obtained of the following documents and drawings for
use during this dam condition assessment.
3.1 Reports and Papers
1) Blueifuss, D.J., Haeke, J.P., Rockfill Dams: “Design and Construction
Problems”, Paper No. 3072, Transactions of the American Society of Civil
Engineers, 1954,.(illustrated with discussions on design and construction of
Kajakai dam).
2) International Engineering Company Inc. “ Feasibility Study for Installation of
Spillway Gates at Kajakai Reservoir, Afghanistan”, August, 1971
3) Ministry of Water and Power , “The Kajakai-Kandahar Power Supply
System”, August 2003.
4) Morrison-Knudsen Afghanistan, Inc., “Final Design report on Kajakai Dam,
Arghandab Dam and Boghra Canal Projects”, prepared for Morrison-Knudson
Afghanistan Inc. by International Engineering Company Inc., December, 1956
5) Harza Overseas Engineering Company, “Contract Documents 912-2, Volume
III, Technical Specifications”, April, 1977

3-2
6) Hitachi Zosen, Hitachi Shipbuilding & Engineering Co., Ltd., “Kajaki Gates
Project, Contract 912-1, Contractors’s Drawings”, February, 1979 (incomplete
copy obtained).
3.2 Drawings
3.2.1 Morrison Knudsen Afghanistan Inc. Drawings
1) 10-F-7: Kajakai Dam Excavation General Plan, September 6, 1951.
2) 11-F-2R: Kajakai Dam Excavation Plan and Foundation Treatment, October
18, 1952.
3) 11-F-1 R2: Kajakai Dam, Plan and Sections, September 28, 1950.
4) Kajaki Dam, Dam Piezometer Installation, unknown drawing number,
unknown date.
3.2.2 International Engineering Company, Inc. Drawings
1) HA-03-205: Kajakai Spillway Gate Study, Existing Spillway Plan and
Sections, April 1, 1971.
2) HA-03-206: Kajakai Spillway Gate Study, Proposed Spillway Arrangement,
April 1, 1971.
3) HA-03-207: Kajakai Spillway Gate Study, Proposed Spillway Crest Details,
April 1, 1971.
4) HA-03-208: Kajakai Spillway Gate Study, Dam, Outlet works & Proposed
Powerplant, April, 1971.
5) GG-12-199: Kajakai Hydroelectric project, Hydrology, July 30, 1971.
3.2.3 Hitachi Shipbuilding and Engineering Co. Ltd.
1) Drawing No. 150997, Kajakai Gates Project, General Arrangement, Sheet
G-1, March 28, 1978.
2) Drawing No. 150998, Kajakai Gates Project, Embedded Parts Assembly,
Sheet G-2, March 28, 1978.

3-3
3.2.4 Water and Power Development Consultancy Services
(India) Ltd.
1) Drawing No. 1011-WAP/KJK-01, Powerhouse and Intake Structure, General
Arrangement, April, 1978.
2) Drawing No. 1011-WAP/KJK-02, Intake Tunnel & Penstocks, Profiles and
Sections, April, 1978.

4 Site Inspection and Condition Assessments

4 Site Inspection and Condition Assessments
4.1 Introduction
A representative of Acres made a site inspection Kajakai Dam and spillways during
February 10 to 14, 2004. Field work included walk-over studies of the dam,
abutments, main spillway and the emergency spillway. While the inspection focused
on the geotechnical aspects of these structures, relevant civil details were also noted.
Limited geological mapping and a discontinuity survey were carried out in order to
define the geotechnical parameters of the bedrock. The results of the field inspection
and geological appraisal are presented in the following sections and on the digital
photographs presented in Appendix A.
4.2 Geology
A geology assessment was made as part of the site inspection and project review
work. This assessment is based a review of published reports and observations made
during the course of the field work.
4.2.1 Kajakai Project Area Description
The Kajaki Dam and reservoir are located in northern Helmand province. The
area is encompasses the southern fringes of the undulating piedmont area which
separates the lowlands of the south from the mountainous terrain in north and
central Afghanistan. The damsite and reservoir area are characterized by
dissected, hilly terrain consisting of sharp featured limestone uplands and rolling,
gentle featured shaly lowlands
The Kajaki damsite is located within a 1 km long, constricted and narrow reach of
the Helmand River. The river valley broadens into a wide basin-like depression in
the areas upstream and downstream of the Kajakai project area.
4.2.2 Regional Geology
The area around the damsite and along the north side of the reservoir consists of
bedrock terrain, either as exposed outcrops or thinly mantled by stony or fine
grained colluvial deposits. Thick colluvial and fine grained alluvial deposits
mantle the mostly shale bedrock in much of the lower lying terrain on the south
side of the reservoir. Widespread granular alluvial deposits occur along the
Helmand valley downstream of the dam and in some of the more significant water
courses north and south of the site.
The regional bedrock geology consists of various formations of Triassic to
Jurassic age carbonates, sandstones, marlstones and shales with occasional

4-2
occurrences of volcanic rocks. The area on the west side of the reservoir,
including the Kajakai damsite, is underlain by a sequence of bedded limestone and
dolomitic limestone.
The bedrock strata have been gently folded into a series of flexures about a
number of northeast trending axes. he bedding is inclined gently towards the
northwest in the terrain on the west and northwest of the reservoir. A number of
faults, mostly aligned in a northwest-southeast direction are present within the
damsite and around the reservoir.
4.2.3 Engineering Geology of the Damsite and Spillway Area
The terrain around the damsite is rugged and generally devoid of significant
vegetation. Exposed rock forms the dam abutments and makes up the major
landforms throughout the project area. The original riverbed of the Helmand
River was covered by several metres of alluvium. Extensive deposits of fine
grained soils mantle the terrain on the left side of the reservoir, southeast of the
dam. Downstream from the dam, granular alluvial deposits gravels occur within
the Helmand River valley and in the low lying ground to the west and north of
Kajakai Bazaar.
The bedrock within the damsite and spillway area consists of a stratified sequence
of limestone and dolomitic limestone. The rock mass consists primarily of
medium thick, gray beds with thin dark gray to black interbeds.
• The medium gray dolomitic limestone is medium to finely crystalline,
generally fresh and strong to very strong (estimated 75 to 150 MPa). It
usually has a uniform and homogeneous texture. The beds of medium gray
rock are generally 40 cm to 150 cm thick and they constitute about 80 to 90
percent of the rock mass.
• The dark gray interbeds are 10 cm to 20 cm thick and are generally spaced
50 cm to 150 cm apart in the rock mass. They vary from fine grained to
medium crystalline. This rock usually has a homogeneous texture but some
beds have very closely spaced laminations which impart a local cleavage.
Outcrop surfaces are usually fresh although a few weathered beds were noted.
The intact rock varies from soft to hard and has a strong compressive strength
(50 to 100 MPa).
The rock mass is generally competent and sound. It contains moderately to
widely spaced bedding planes and widely to very widely spaced joints, which
impart a tabular blocky characteristic. Three major sets of discontinuities have
been identified. Other random joints are present and some of these may form

4-3
subsets of the major sets. Results of the discontinuity survey, which consisted of
60 measurements at the dam abutments and in the spillway channels, are shown
on the stereographic plot on Figure 5.
The major discontinuity sets are described in Table 4.1.
Table 4.1
Description of Major Discontinuity Sets
in the Kajakai Dam and Spillway Areas
Dip
Dir./Dip Spacing
Set (deg) (m)
Bedding 263/11 0.2 to
0.8
J1
(joint)
J2
(joints
and
faults)
J3
(joint)
152/88 2.0 to
10.0
244/84 4.0 to
10.0
128/85 2.0 to
>10.0
Persistence Surface
(m) Characteristics
> 50 Smooth to
rough, planar
10 to >50 Very rough,
irregular to
planer
frequently
weathered
5 to 15
(Faults are
> 50 m)
Very rough,
irregular to
planar,
frequently
weathered.
30 to > 50 Very rough,
irregular to
planer
frequently
weathered
Description
Planar, continuous
surfaces, some karst
solutioning evident in a
number of features
Major joint set.
Frequently affected by
karstic solutioning, with 1
cm to 30 cm aperture.
Usually infilled with
plastic clay and/or rock
debris.
Secondary joint set which
has a more random
distribution than J1. A
few significant faults are
parallel to the J2 joint
system. Occasional
karstic solutioning and
clay/and rock debris
infilling
Poorly defined secondary
joint set which overlaps
with set J1. Frequently
affected by karstic
solutioning, with 1 cm to
30 cm aperture. Usually
infilled with plastic clay
and/or rock debris
The bedding planes and joints of set J1 make up the dominant fabric of the project
rock mass. Individual bedding planes are usually continuous and numerous
master joints of set J1 have 50 to more than 100 m continuity. Joints of this set
include a large number of prominent, clay infilled, karstic solutioned “Karen” type
features which run across the dam and spillway area. The joints of set J2, which
are aligned normal to J1, usually have 5 to 20 m continuity but a few features are
up to 50-m long. The persistence of J3 joints is similar to J1 discontinuities.

Figure 5
Kajakai Dam Condition Assessment
Kajakai Dam and Spillway Discontinuity Survey

4-5
Most bedding surfaces are locally planar and rough. Joints vary from planar to
wavy and are usually rough. Joints are usually open near the ground surface but
tend to close with depth. Clay coatings and infillings of up to 30 cm thick occur
in most J1 and J3 joints and in some J2 joints. Most joints exhibit some calcite
lining and hematite staining. Calcite infillings are present in some discontinuities.
Clay filled joints have low shear strength and require special attention when
considering stability and water seepage issues.
Bedding planes dip gently in a downstream direction. These features, which
dominate the structural geology regime, are unfavorably oriented for the sliding
stability of water retaining gravity structures. The J1 and J2 joint sets are oriented
obliquely to the dam axis. The J1 joint set cuts across the axis of the main
spillway channel while the J2 set is aligned almost parallel to it.
The rock mass has been affected by karstic solutioning. This phenomenon is
believed to be most prevalent in the stress relieved/ weathered zone within about
30 to 40-m from the ground surface, but it can be encountered at greater depths
within the rock mass. Karstic alteration of the rock mass is evidenced by the
following features:
• Solution enlargement of discontinuities. Solutioned joints generally have 1
to about 30 cm aperture. These features are usually infilled with clay/silt
although a number of features were observed to be open. Rock debris
infillings were noted at a few locations. Karstic solutioning has affected most
joints of the J1 set. In addition to the J1 joints, numerous bedding planes, J2
joints and a few random joints are affected by solutioning.
• Solution vugs and cavities. Most solution cavities are less than 30 cm
diameter but a few features of up to 150 cm diameter were noted.
• Partial decalcification of a few thin dolomitic limestone layers. This effect
leaves a porous, relatively weak rock with a porous Achalky@ texture.
4.3 Dam
The 90 m high embankment is generally in good condition. A comprehensive
walkover survey covered the crest, each berm on the upstream and downstream faces,
and the four abutment groin areas. The riverbanks downstream of the dam were
examined for signs of water seepage. Observations made during the field inspections
are summarized in the following sections.

4.3.1 Crest
4-6
The el 1050 m crest is a roadway which provides access to the right bank,
including the spillways (Photo D3). The crest is level and the roadway is well
maintained with no significant rutting or pot holes. There were no signs of
erosion or settlement along the crest.
Prior to the site visit, three test pits were excavated in the dam crest by Kajakai
powerhouse staff. These test pits were located downstream of the crest centreline
(Photo D3) and were 3.0 m deep each. The purpose of the test pits was to
determine the properties of the upper levels of the impervious core. The three pits
were located at the following approximate locations.
• Test Pit No. 1: approximately 60 m from the right abutment
• Test Pit No. 2: approximately 127 m from the right abutment
• Test Pit No. 2: approximately 190 m from the right abutment.
All three test pits encountered similar materials. A general description is as
follows:
• The crest is covered by 0.70 m to 0.95 m of road bed material. This consists
of a densely compacted sand and gravel. The material well graded and
contains trace amounts of silt and clay.
• In Test Pit Nos. 1 and 2, the road bed material is underlain by a 1.0 m thick
horizontal layer of brown, moist, densely compacted sand and gravel with
trace amounts of silt and clay. This material, which is believed to be part of
the “compacted free draining granular” zone, is absent in Test Pit No. 3.
• Compacted rockfill is present in the downstream half of the lower 1.5 m of all
three test pits. The rock fill consists of dry, well graded, angular particles
which range in size from about 5 mm to 100 mm. It contains some sandy
material in the interstices.
• Compacted silty sand with some gravel occurs in the upstream half of all three
test pits below about 1.5 to 2.0 m depth. This very dense, moist material is
coarser than the “impervious” fill material described Table 2.4. It could be the
so called “semi pervious fill” which may have been used in the filter zone
immediately downstream of the impervious core.
The test pit program was inconclusive, since the impervious core could not be
positively identified. The test pits did show, however that the fill material is very
well compacted with no signs of cracking or other unfavorable properties.

4-7
Samples were taken from all three test pits and these have been submitted for
index properties testing. Test results were not available at the time of writing.
4.3.2 Upstream Face
Two berms are visible on the upstream face of the dam, the el 1040 m and el
1030 m berms. Observations of the upstream face are summarized as follows:
• The reservoir level was at approximately el 1029.5 m, just below the level of
the el 1030 m berm, at the time of the inspection on February 11 to 13. The
high water marks visible on the rockfill slope were estimated to be at about el
1036 m.
• There are a number of loose boulders on the inside half of the el 1040 m berm
surface (Photo D4).
• The el 1030 m berm is littered with medium to large boulders, which have
rolled down from the embankment slope (Photo D5). There are some signs of
minor beaching type erosion along the inside edge of this berm. This has
resulted in localized over-steepening of the toe of the el 1030 to el 1040 interberm
slope.
• The el 1030 m to el 1040 m inter-berm slope is very steep (as per the design
drawing showing on Figure 4), probably only slightly flatter than the angle of
repose. A number of shallow scalloped shaped sloughs are visible along the el
1040 m crest-line and the upper portion of the inter-berm slope. The slip
depressions extend 1 to 2 m into the crest of the berm. Some of these features
are related to toe over-steepening, caused by the el 1030 m beaching erosion
described above. The main areas of crest edge loss are at the following
approximate locations.
o Sta. 0+290 to Sta. 0+ 305
o Sta. 0+310 to Sta. 0+ 325
o Sta. 0+ 340 to Sta. 0+ 355
The el 1040 crest line between approximately Sta. 0+185 and Sta. 0+250 is
irregular with variable upper slope erosion.
• The riprap slope protection, while loosened and requiring some dressing is
effective below approximately el 1045 m. The riprap on both inter-berm
slopes has suffered particle segregation. Finer rockfill (20 to 100 cm block
size) predominates the upper levels of each slope. The lower levels have
much coarser, loosened rock fill and boulders (up to 1.5 m block size at el

4-8
1030 m). Particle interlocking is poor, particularly on the lower slopes, just
above el 1030 m and el 1040 m. (Photo no. D4)
• There is a slightly concave slope break midway up the el 1040 to 1050 m face.
It appears that a finer rockfill was placed on the crest of the dam at a later date
and this material covers the slope above approximately 1045 m. The upper
half of the el 1040 m to el 1050 m face is slightly steeper than the lower half.
4.3.3 Downstream Face
The overall downstream face is shown on Photos D7 and D8. It consists of a total
of 8 berms, which are spaced 10 m apart. While the overall face has a slope of
2:1, the inter-berm slopes are much steeper, generally in the range of 1.3:1 to
1.4:1. The face is covered by loose rockfill. In general, there is no evidence of
significant erosion or settlement, except for the few cases mentioned in the
following paragraphs.
The following observations were made during the walk-over inspection of the
downstream face of the dam:
• In general, the rockfill on each 10 m high inter-berm slope shows particle
segregation, with finer rockfill at the top and coarser rockfill at the base and
toe.
• The berms and slope faces are generally in very good condition. The berm
surfaces are generally clear although a number of boulders, which have rolled
down from the slope, were seen at numerous locations.
• The rockfill visible on the dam face below the el 989 m berm is significantly
coarser than that above this level. (Photos D7, D8 and D10). According to
Harza Drawing11-F-1 R2, Kajakai Dam, Plan and Sections, this section of the
dam incorporated the original downstream cofferdam, which was placed much
earlier the main body of the embankment. The face of the dam in this lower
section is covered by boulders that are up to 1.5 m in diameter.
• A small slough type slide was noted in relatively fine grained rock fill at the
left (south east) end of the el 1040 m berm. The slough is shallow and is
confined to the upper half of the el 1030 m to 1040 m inter-berm slope. This
feature could be easily repaired by some slope dressing with a backhoe.
• There is a small crater along the el 1030 m crest-line, about 5 m from the right
abutment of the dam (Photo D9). This crater is about 1.2 m deep and

4-9
approximately 2.8 m to 3.0 m in diameter. Kajakai staff reported that it was
caused by a bomb explosion.
• A small hole was noted in the middle of the dam on the el 980 berm (Photos
D11 and D12). The origin of this hole is unclear. It does not have the
characteristics of a bomb blast crater but appears to be related to some
settlement or shifting of the underlying rockfill. As can be seen on the
photographs, a 0.5 m diameter cavity is visible at the southwest side of the
hole. This cavity is overlain by a rockfill “bridge” that is about 0.5 m thick.
There are some debris around the lip of the hole, which suggest that some
hand excavation has been carried out to enlarge it at some time in the past.
Two bench marks are located about 3 m and 10 m from the hole. One of the
bench marks has the inscription “EDGE WALL”, suggesting that some
concrete structure may be buried in the rockfill. There are three possible
explanations of the origin of the hole in the el 980 berm:
o Bomb crater (highly unlikely, given the shape of the hole)
o Piping of fill beneath the berm as a result of heavy water flows when
sluicing the rock fill during dam construction or seepage through the
dam core after impoundment
o Collapse or shifting of a buried concrete structure.
4.3.4 Seepage
There is limited evidence of seepage through the dam or its foundations.
Powerhouse staff report that there was some seepage from the toe of the dam
during the early 1950s but this area was covered by granular fill material more
than 45 years ago and there were no signs of seepage in the dam toe at the time of
the site inspection.
Seepage could be seen at the base of the cliff in the downstream right bank area.
This seepage consists of the following:
• Vegetation in the lower right groin of the dam between el 970 and 975 m
(Photo D13). This could be an indication of damp soil caused by minor
seepage in that area.
• Seepage from bedding planes 2.0 m to 3.0 m above the tailpond level in the
right bank cliff downstream of the dam (Photos D14 and D15).
No seepage could be seen in the downstream left abutment or in the left groin of
the dam. A few small seepages were seen in the left bank tunnels, just
downstream of the plugs.

4-10
4.3.5 Embankment Instrumentation
An instrumentation cabinet is located at the base of the cliff, downstream from the
toe of the dam (Photos D16 and D17). The cabinet, Bourdon gauges and plastic
tubing are all in good condition but it is not known if the system is functional.
The gauges are connected to the piezometers, which were installed in the
embankment dam during the early 1950s. The condition of the piezometers in the
dam is unknown. The Kajakai staff report that no piezometer readings have been
made during the past 25 years or so. Staff had to break the lock to the cabinet in
order to gain access during the dam inspection.
4.4 Service Spillway
Inspections were carried out of the service spillway. Copies of digital photographs
taken during this inspection are shown in Appendix A. Observations made during the
inspection are given in the following paragraphs.
4.4.1 Concrete Structure
Aerial views of the spillway and the concrete structure are shown on Photos S1
and S2. Walkover inspections were made of the partially completed concrete
structure and the following were noted. The “as built” structure has a significantly
different design from that shown on the 1971 International Engineering Company,
Inc. Drawing HA-03-206: Kajakai Spillway Gate Study, Proposed Spillway
Arrangement. Significant differences between the drawing and the as-built
structures include more steeply inclined rollway surfaces, concrete lined spilling
basin excavated into the spillway floor, the provision of a drainage gallery and
possibly a different site location.
Field observations include the following:
• The crest of the concrete structure (Photo S3) is estimated to be in the range of
el 1035 to 1035.5 m, as compared to the design level of el 1037 m for the final
ogee sill (Hitachi Shipbuilding and Engineering Co. Ltd Drawing No.
150997).
• Wing walls and sloping concrete rollway surfaces (Photos S4 and S5) have
been partially completed for six of the planned eight bays.
• The concrete is generally in good condition and little preparatory work would
be needed to incorporate existing works into a final structure.

4-11
• Considerable embedded rebar is visible (Photos S3, S4, and S5) in all concrete
structures. The exposed rebars, while showing some corrosion, (Photo S6) are
generally in fair condition. Many bars have been bent by spillway operation
during the more than 25 years since the structure has been constructed.
• The stilling basin has been excavated about 3 m below the floor of the
downstream spillway (see Photo S7). The concrete lining is mostly
completed.
• A 15 m wide concrete apron slab has been placed on the floor of the spillway,
just upstream of the rollway structure.
4.4.2 Geotechnical Aspects of the Spillway
The spillway is excavated in dolomitic limestone bedrock, which is described
above in Section 4.2. The bedding is inclined about 10 to 15 degrees in a
downstream direction. J1 joints are visible in the rock slopes on both ends of the
structure. A fault zone runs along the spillway channel, parallel to the J2 joint set.
Observations of the rock foundation conditions are summarized as follows.
• Contrary to some earlier reports, no rock anchors were installed in the
abutments of the gate structure. (A few dowels were installed loosely into a
pattern of holes in the upper left bank, well downstream of the concrete
structure. The purpose of these dowels is unclear although they may have
been installed to support some equipment or apparatus associated with the
earlier construction work).
• As seen on Photos S8 and S11, subvertical J1 joints are present in each
abutment upstream and downstream of the concrete foundation structure. The
ends of the concrete structure abut against relatively unjointed rock. However,
the J1 joints are aligned oblique to the structure axis and many cross the
foundation some distance from the abutments.
• The J1 joints have been effected by karstic solutioning(Photos S9 and S 10).
This solutioning extends the full height of the rock slopes and likely extend
some depth below the base level of the spillway. Most of the J1 features are
infilled with clay and/or rock debris. In many cases, the clay infilling has been
washed out to a depth of a few metres below the excavated rock surface
(Photo S10) since 1980. This has resulted in the rock mass having a loosened
unstable appearance in a few locations. (Photo S9).

4-12
• The 20 m to 40 m high rock slopes are generally stable against deep seated
sliding failure. Blasting of these slopes was fair to good.
• There are a number of locations where weathering and/or joint conditions
result in shallow seated rock-fall hazards in the rock slopes. Some of the more
notable locations where rock fall hazards may require further attention are :
o Right bank area with closely space J1 joints immediately downstream of
the right abutment of the concrete rollway structure (Photo S9), where
the loosened rock blocks constitute a rock fall hazard.
o Left bank area, in the concrete rollway and stilling basin area. Numerous
loosened blocks need scaling or rock anchor support.
o In the right bank, some of the J2 joints are inclined 70 to 75 degrees
outwards, undercutting portions of the slope. These may trigger small
scale block sliding in the upper slope above the concrete structures.
Further studies should be made to assess this hazard.
• A gallery has been excavated into the right bank from the base of the stilling
basin (Photos S13 and S14). The portal excavation slopes downward and
extends about 5 m into the rock. The gallery branches in upstream and
downstream directions from the end of the portal gallery. The gallery was half
filled with water and could not be entered at the time of the inspection. This is
believed to be a drainage gallery.
• The fault zone in the centre of the spillway channel (Photos S15 and S16)
contains about 1.5 m to 2.0 m of unconsolidated gouge material and is
associated with some significant karstic solutioning. The fault is located
immediately upstream of the fifth spillway bay from the right abutment.
Extensive dental excavation and concrete backfilling work was apparently in
progress at the time that work was suspended in 1980 (Photos S17 and S18).
Some details of this work, as seen during the site inspection are as follows:
o The excavation is 2.0 to 1.5 m wide and extends from 4 to 28 m upstream
from the upstream edge of the concrete rollway structure. A cutout has
been left in the upstream apron to accommodate this excavation. The
excavation had a maximum depth of about 3 m at the time of the site
inspection.
o The downstream end of the fault zone excavation intercepted a karstic
cavity. This cavity was seen to be up to 2 m wide. It developed along a
J1 joint, which intersects the fault at a right angle. It could be seen to
extend more than 4 m away from the fault zone. The cavity was mostly
infilled with water and could not be explored at the time of the site
inspection.

4-13
o The floor of the fault excavation is covered in debris and it is not clear
how deep the original dental excavation was.
o Concrete backfill was seen on the downstream wall of the karstic cavity
(Photos S17 and S18). The extent and depth of the backfill could not be
determined.
4.5 Emergency Spillway
The emergency spillway is uncompleted and is not excavated to final grade. This
facility has no present impact on the operation of the reservoir.
It is evident that work was halted quickly in 1980. The blasted rock muck piles and
excavation equipment are still in place where they were abandoned almost 24 years
ago (Photo S20) No plans or designs of the emergency spillway were available at the
time of the inspection and it was difficult to determine how much excavation work
needs to be done. It is understood that this facility was intended to include a low
lying fuse plug embankment water retaining structure.
The geology is similar to that encountered in the nearby main spillway. The clay
filled J1 and J2 joints crisscross the site. These features have to be dealt with when
constructing any water retaining fuse plug structure in this area.
It is noted that there is an unusual amount of live ammunition and other unexploded
ordinance scattered in the emergency spillway. This constitutes a very serious safety
hazard.
4.6 Water Conveyance Structures
Brief examinations were made of the penstocks and the mechanical components of the
power and irrigation facilities. All equipment was seen to be in generally fair
working order except for the upstream irrigation outlet valve, which was stuck
partially open at the time of the inspection.
Both tunnels were entered and inspected. The tunnel excavations are located in
excellent quality rock and are unsupported. The tunnels are unlined. The quality of
the tunnel excavations and associated portal surface excavations is remarkably good.
Most control perimeter blast hole traces are visible and there is very little overbreak.
No signs of rock instability were noted. There is some minor seepage from rock
joints in the vicinity of the power tunnel plug, particularly in the right haunch and
wall. Very little seepage could be seen in the irrigation tunnel.

4.7 Reservoir
4-14
The reservoir, which is shown on Figure 2, extends along a 26 km long reach of the
Helmand River. It is divided into two bodies of water by a 300 m wide rocky narrows
(Photos R1 and R2), which is located approximately at the mid point of the lake, some
15 km upstream from the dam. These sub-reservoirs are designated Upstream
Reservoir and Downstream Reservoir on Figure 2. The Upstream Reservoir is
approximately 11 km long and 2 to 4 km wide (Photos R3, R4 and R5). The
downstream section of the reservoir is about 15 km long and approximately 1.5 to 5
km wide (Photos R6, R7, R8 and R9). While no systematic study of the reservoir was
carried out, a number of observations were made during helicopter flyovers.
• Rocky uplands make up most of the northwest, north and northeast shores of the
Upstream Reservoir and the northwest and north shores of the Downstream
Reservoir. This terrain consists of gently inclined limestone strata with very little
soil cover.
• Low lying, subdued topography forms much of the south shoreline of the
Downstream Reservoir (Photo R8). This terrain appears to be underlain by a soft
shaly rock formation and is mantled by a continuous cover of fine grained soil.
• A large alluvial fan juts into the east side of the Upstream Reservoir and
constitutes about half of the eastern shoreline (Photos R4 and R5). This flat lying
landform has very low relief and hosts some agriculture.
• While small land slips cannot be ruled out (Photo R5), no evidence of large
unstable rock masses or land slides could be seen. It appears that the annual 20 to
35 m of reservoir level fluctuation has not triggered any large scale slope
instability in the reservoir.
• A plume of cloudy water could be observed in the Upstream Reservoir (Photo
R3). The zone of cloudy water extends along the entire length of the northwest
side of this section of the reservoir. It extends through the Rocky Narrows
(Photos R1 and R2) and dies out in the upstream end of the Downstream
Reservoir. The origin of the cloudy water is uncertain. When viewed from an
aircraft, it resembles an algae plume and is superficially similar to the one which
occurred in Lake Mead in the USA during 2003. It is also possible that the
cloudiness is indicative of suspended silt in the water.
• Brown algae masses were observed floating in the water near the dam and in the
rocky narrows (Photo R2).

4-15
• The reservoir area is very sparsely inhabited. There are a few villages near the
shorelines of both the Upstream Reservoir (Photo R4) and Downstream Reservoir
(Photo R9). Farming cultivation extends down to the water’s edge in the alluvial
fan on the east side of the upstream reservoir. It is not clear if any dwellings are
close enough to the shore line to be affected by flooding when the spillway gates
are installed.

5 Assessment of Dam

5 Assessment of Dam
5.1 Embankment Dam Stability
5.1.1 Stability Analyses
Slope stability analyses were carried out of the Kajakai Dam using RocScience
SLIDE software. SLIDE is a 2D slope stability program for evaluating the safety
factor of circular or noncircular failure surfaces in soil or rock slopes. It analyzes
the stability of slip surfaces using vertical slice limit equilibrium methods. The
stability analyses carried out on the dam used the critical surface search technique
in order to locate the critical slip surface for each given slope. Two analytical
methods, Morgenstern-Price and Bishop, were used.
Three loading cases were analyzed.
• Normal
• Rapid drawdown
• Seismic (pseudostatic).
The results of the analyses are shown in Appendix C and are described in the
following paragraphs.
5.1.2 Geotechnical Parameters
In the absence of site specific test data and construction records soil and rock
properties were assumed in the analyses as being representative of actual
conditions. These properties are shown in Table 5.1.
Table 5.1
Representative Materials Properties Assumed
for Stability Analyses of the Kajakai Dam
Material Name Density
(KN/m 3 Friction Angle Cohesion
) (deg) (KPa)
Sluiced Rockfill 18.5 40 0
Impervious Rolled Fill 19.5 30 0
Transition Zone 19.5 37 0
Compacted Free Draining
gravel
19.0 40 0
Riverbed Granular
Alluvium
21.0 38 0
Limestone Bedrock 27.0 55 100

5.1.3 Reservoir Levels
5-2
The stability analyses were carried out using the original design reservoir and
tailwater levels. These are given in Table 5.2.
Table 5.2
Reservoir and Tailwater Level Used for
Stability Analyses of the Kajakai Dam
Analysis Case
Elevation (m)
Reservoir Tailwater
Normal 1045 978
Rapid Drawdown 1037 978
Seismic 1045 978
5.1.4 Seismic Parameters
The 1956 Morrison Knudson design report indicates that an earthquake loading of
0.05 g was used in the original design of the Kajakai dam. This seismic
coefficient was applied as pseudostatic loading in Felenius type sliding block
stability analyses of the dam embankment. It was assumed that this earthquake
ground motion would be the result of seismic activity in the high hazard zone in
Pakistan about 320 km form the damsite. It was also assumed that there was no
significant local seismic activity.
The Kajakai area is now considered to be more seismically active than had been
earlier assumed. The Global Seismic Hazard Assessment Program (GSHAP)
seismicity map of Afghanistan and Pakistan (available on the GHSHAP website)
has classified the Kajakai area as a zone of moderate seismic hazard. A
preliminary seismicity study has been carried out for the purposes of this dam
condition study. The seismicity study is described in Appendix B.
In view of the preliminary nature of the seismic studies, dam stability was checked
for a range of seismic coefficients. Two levels of earthquake ground motion have
been considered.
• Operating Basis earthquake (OBE)
• Maximum Design Earthquake (MDE).
The OBE is adapted from the GSHAP. The GSHAP seismic hazard assessment of
Afghanistan is shown on Figure B2 in Appendix B. This map shows the ranges of
Peak Ground acceleration (PGA) values for a return period of 475 years. This

5-3
level of earthquake is appropriate for buildings and most civil structures but is not
appropriate for determining the overall stability of a high dam such as Kajakai.
The MDE is based on a review of earthquake records in the region and a
preliminary assessment of the Maximum Credible Earthquake. This data and
MCE estimation are discussed in Appendix B. The MCE is the largest earthquake
which can be reasonably expected to occur in an area, given its geology, tectonics
and earthquake history. The MCE events selected herein are preliminary criteria,
which are subject to revision when more detailed seismicity studies are carried
out. The preliminary assessment indicates that two MCE events should be
considered when calculating the MDE for design.
• Near field Magnitude (M) 5.5 earthquake, which has a hypocentral distance of
15 km from the damsite. This is considered to be the maximum earthquake
for the “moderate risk” area surrounding the damsite.
• Far field M 8 event, which occurs 150 km from the damsite. It is assumed that
this earthquake is generated in the high hazard zone located east and southeast
of Kandahar.
Table 5.3
Peak Ground Acceleration and
Sustained Seismic Coefficient Values
Earthquake
Operating Basis
Earthquake (OBE)
(1/ 475 year event)
Maximum Design
Earthquake (MDE)
Peak Ground
Acceleration
(PGA, g)
Seismic coefficient, k,
for Pseudostatic
Analyses
(2/3 x PGA)
0.16 0.11
0.25 0.17
Both the MDE and OBE have been used in the stability analyses. The OBE gives
a stability check using the lower range of expected seismic events. The MDE
earthquake permits the most conservative estimate of the effects of earthquake
loadings. The PGA value of 0.25 g for the MDE event is equal to or greater than
the maximum ground motion caused by either a near field M 5.5 near field
earthquake or a far field M 8.0 event. The return period of this PGA is not known,
but is expected to be in the range of 2000 to 8000 years. A site specific
probabilistic analysis should be carried out at some future date in order to
determine the ground motion return periods.

5.1.5 Analysis Results
5-4
The stability analysis results are presented in the Table 5.4. Note that the results
of the Bishop method are shown below. The Morgenstern-Price results, which are
very similar, are shown on the figures in Appendix C.
Table 5.4
Kajakai Dam Stability Analysis Results
Analysis Case Upstream Face
Factor of Safety
Downstream Face Minimum FOS( 1 )
Normal Loading 2.20 1.75 1.5
Rapid Drawdown 2.04 na 1.2
Seismic 0.11 g 1.29 1.31
Loading 0.17 g 1.03 1.14
1.1
(1) Federal Energy Regulatory Commission (FERC) “Engineering Guidelines for the Evaluation
of Hydropower sites”, April 1991 (updated in November 2003).
The analyses indicate the following:
• The FOS values for the normal and rapid drawdown static loadings are
satisfactory.
• The FOS is adequate for the DBE and MDE seismic coefficients of 0.11g. and
0.17 g respectively.
The relatively low upstream FOS for the MDE indicates that, while the dam is
stable, relatively high deformations may occur during a severe earthquake. More
sophisticated analyses should be carried out to determine if earthquake induced
deformations are within acceptable limits.
5.2 Seepage and Drainage Control
As reported in Section 4.3.4, very limited seepage flows were evident in the
downstream area of the dam during the field inspection. The dam is built on slightly
karstic limestone. This type of rock mass is often prone to seepage problems when
present in the foundations of large dams. Clay filled discontinuities, similar to the
ones known to be present at Kajakai, are usually susceptible to long term wash-out
and erosion. Despite the nature of the foundation geology, the site inspection
indicates that these problems have not developed during the 50 years of operation of

5-5
the Kajakai dam. It should be noted that the reservoir was never impounded to design
levels during this period.
The observed seepage downstream of the dam is very low and is generally much less
than is normally seen in the limestone foundations of large dams. As described in
Section 4.3.4, there is a line of seepage near the toe of the cliff downstream of the
right abutment of the dam. No seepage could be seen in the downstream left
abutment area. A few small seepages were seen in the left bank tunnels, just
downstream of the plugs.
The low seepage rates are evidence of the high quality of the grout curtain. A brief
review of some of the 1952-53 field correspondence, which was found in old files at
the site, indicates that a well executed program of grouting was carried out. Detailed
records were kept and design modifications were implemented to cope with
unexpected field conditions as they were encountered. References were made to
borehole washing operations to remove clay seams in the foundation, prior to the
injection of grout. The designers also carried out extensive dental excavations and
concrete backfill works in faults and karst zones (see Section 2.4.3 above). These
works show that a very conscientious effort was made to remove erodible material
from the foundation bedrock of the dam.
There is one area of concern, which should be investigated further. The powerhouse
operators reported that there was seepage from the toe of the dam shortly after
impoundment in the early 1950s. It is possible that this occurred in the area of the
buried channel which was encountered in the old riverbed (see Section 2.4.3 above).
Remedial works, including the placement of granular backfill at the toe of the el 970
berm, were reportedly carried out to deal with this seepage in the 1950’s. The
backfilled toe of the dam is now at about el 967 m and no seepage can be seen. It is
noteworthy that there is an unexplained hole and cavity beneath the surface of the el
980 m berm (see Section 4.3.3 and Photos D11 and D12). This feature is located
roughly in the area where the old seepage was reported. While there are alternative
explanations for the origin of this cavity, the possibility of seepage related sinkhole
development cannot be discounted. This feature should be studied further.
5.3 Deformation and Cracking
No visible signs of settlement or cracking could be seen during the site inspection of
the dam. Without a survey, it is not possible to determine if the original longitudinal
crest camber is still present or if the dam body experienced significant settlement in
the years following construction.
As discussed earlier, it is not clear if the dam crest was raised above the original el
1050 m level during the1979-80 construction program. This must be determined by

5-6
surveying and an examination of construction records, before any program of
monitoring can be implemented.
5.4 Surface Erosion
A limited program of remedial works is needed to repair areas of surface erosion in
the upstream face and in a few local areas of the downstream face. As discussed, in
Section 4.3.2, there has been minor beaching erosion and localized slumping of the
berm slopes in the upstream face. A few small slumps were also seen the downstream
face. None of these features endanger the stability of the dam. Features of this type
are typical of an old dam and they should be repaired as part of a normal dam
maintenance program. Slope dressing is needed on the upstream face to remediate the
rip rap slope protection and to halt the process of shallow berm slope slumping.
5.5 Liquefaction Potential
Loose, cohesionless soils tend to densify when subjected to earthquake induced
ground vibrations. If the soils have limited drainage capability, drainage will be too
slow to allow the soil to change volume quickly, and pore pressures will build up
rapidly. This buildup in pore pressure reduces the effective stress and hence the
available shear strength of the material. This phenomenon is known as liquefaction.
It normally affects saturated, loose to medium compact sand but has been to known to
occur in gravels and some rockfill material.
Dams, which have apparent static stability against earthquake motions, may be
susceptible to liquefaction damage or failure. Liquefaction potential of an
embankment dam or its foundation is a function of the severity of ground motion,
particle size distribution of the material, groundwater conditions, overall permeability,
and the degree of compaction (density index) of the fill material. Embankment dams,
which are constructed of well compacted fill materials, generally have a low
vulnerability to liquefaction.
The Kajakai Dam has a few of factors which indicate that a liquefaction potential
assessment should be carried out.
• The assumed PGA of the MDE is 0.25 g. This level of ground motion is well
within the range that is known to cause liquefaction.
• The upstream and downstream shells of the dam were constructed on the in-situ
riverbed alluvium. This stratum is about 3 m thick and extends along the base of
the dam from the edges of the core to the upstream and downstream toes. It
consists of a granular alluvial deposit which is believed to have a relatively low

5-7
degree of compaction. Submerged alluvial deposits are often susceptible to
liquefaction.
• The upstream and downstream rockfill shells are constructed of sluiced, dumped
material and therefore have not been systematically compacted. The uncompacted
shells can be expected to experience compaction and deformation during an
earthquake. Depending upon the particle gradation and overall permeability, there
may be some potential for liquefaction in the submerged section of the upstream
shell. It must be noted, however, that rockfill generally has high permeability
which enables the rapid dissipation of dynamically induced pore pressures. The
permeability is a direct function of the percentage of finer material present. The
rockfill also consists of interlocking angular particles which have relatively high
shear strength.
The central body of the dam, including the core, filter and granular transition zones, is
made up of compacted fill. Assuming that the material was adequately compacted, it
has a low liquefaction potential.
The Kajakai Dam is considered to have a relatively low susceptibility to liquefaction.
As can be seen, however, there is some question about the liquefaction resistance of
the natural alluvial material beneath the base of the dam and to a lesser extent of the
submerged upstream rockfill shell. It is concluded that future studies are needed to
conclusively assess the liquefaction potential of the dam and its foundation.
5.6 Instrumentation
The dam was instrumented with 13 piezometers at the time of its construction. The
remote reading facility is in good condition but it is not known if any of this 50 year
old system of buried piezometer tips and tubes is still functional. No instrumentation
readings have been made for more than 25 years.
It would be beneficial to be able to monitor pore pressures and embankment
deformations when the spillway gates are installed and the reservoir is raised to its
design level. The existing piezometer system should be evaluated and tested by an
experienced instrumentation engineer to determine if any of the instruments are
functional.
No evidence of a surface movement monitoring system can be seen at the dam. A
network of surface monuments should be installed in order to monitor any
deformations of the crest and slopes, which may result from raising the reservoir.

6 Spillway Assessment

6 Spillway Assessment
The review of the spillway focused on the engineering geology and geotechnical
aspects of the gates structure foundation. A walkover inspection was made of the
concrete structure, but a detailed assessment was not made of the concrete works.
6.1 Concrete Structure
The concrete structures are generally are in good condition. The embedded rebars are
likewise in generally good condition but there has been some corrosion and damage
over the years. Further evaluation of the existing structure would require core
sampling and laboratory testing of the concrete and a further studies and assessment
of the rebars in conjunction with a review of the design drawings.
No engineering drawings of the present concrete structure were available for the
current condition assessment. These must be made available to permit a meaningful
engineering assessment and enable designs for future work to be finalized.
6.2 Slope Stability
The high rock slopes on both sides of the spillway channel have suffered some
degradation since they were excavated more than 50 years ago. As described in
Section 4.4.2, there are numerous clay filled joints and occasional zones of weathered
rock in many places. These have suffered additional weathering as well as general
localized loosening of rock blocks. The slope has a moderate rock-fall hazard in
many places, including areas where work will be carried out when construction of the
gate structure recommences. Scaling and localized rock support work must be carried
out in the work areas as a first step in the construction program.
6.3 Seepage Control
The limestone rock mass is slightly to moderately karstic. Overall permeability is
high and seepage prevention measures, including the construction of a grout curtain
and drainage measures, must be implemented. Special measures to ensure proper
seepage control must include the removal of erodible clay and fault gouge infillings
and replacing them with concrete cut-offs or grout. It is known that these measures
were part of the construction design which was being implemented before
construction work ceased in 1980.
A number of items of geotechnical background information must be established
before the final evaluation of spillway seepage control measures can be made. These
include:

6-2
• as-built details of grouting work carried out in the foundation along the structure
axis
• design drawings of the planned grout curtain and drainage works
• geological mapping, drilling records (if any) and in-situ surveys of the karstic
fault, which is located on the upstream side of the concrete structure, in the
centre of the spillway channel
• as-built details of dental excavation and replacement concrete in the karstic fault
zone adjacent to and under the concrete structure
• details of the design and as-built works of the right bank drainage tunnel.
6.4 Foundation Shear Strength
A concrete structure must be designed to resist sliding on its bedrock foundation. A
sliding stability analyses would normally be carried out as part of a condition
assessment of this type of structure. This could not be done at the present time
because of the lack of design drawings and as-built details.
The rock mass is very competent and has an adequate bearing capacity for the type of
structure planned. The bedding, however, dips about 10 degrees in a downstream
direction and is unfavorably oriented for sliding stability of the gate structure. If this
results in stability issues it can be remediated by measures such as a foundation key
excavation and/or the possible use of rock anchors. An understanding of the
concrete/rock interface shear strength and the bedding plane shear strength is
therefore required to allow a design assessment to be made.
6.5 Outlet Channel Excavation
The original spillway design, as shown on Drawing No. 10-F-7: Kajakai Dam
Excavation General Plan, included a channel shaping excavation at the outlet of the
natural discharge channel into the Helmand River. The purpose of this excavation
was to reduce or prevent flow from the spillway from infringing on the powerhouse
tailrace. This excavation was never carried out, presumably because the spillway
gates were never completed. A hydraulic engineering review should be carried out to
determine if this excavation is needed when the spillway gates are installed.
6.6 Emergency Spillway
It is understood that the incomplete excavation between the right abutment of the dam
and the main spillway is the beginnings of the emergency spillway channel. No
details of the design of this structure are available. Hydraulic and civil design details
are needed to carry out a review of the emergency spillway. Alternatively, a
reanalysis of the reservoir operation in general in the context of the existing spillway

6-3
capacity and a re-examination of the overall flood hydrology and handling capability
will be needed in order to redo the emergency spillway design.

7 Summary and Recommendations

7 Summary and Recommendations
7.1 Summary of Findings
The following paragraphs summarize the condition of the Kajakai Dam and related
structures. It is, however, recommended that the actions recommended in Section 7.2
be taken prior to the initiation of any changes to the present operating circumstances
of the dam.
Dam
• The dam is in good condition. It is possible that there has been some partial
desiccation of the upper levels of the core above the past impoundment levels, but
this is not a significant impediment to raising the reservoir level. The reservoir
can be safely filled to the design impoundment level of el 1045 m, provided that
the dam is effectively monitored and the initial filling is carried out a carefully
controlled rate.
• Slope stability analyses of the dam embankment indicate that it is acceptably
stable and will withstand the effects of full impoundment of the reservoir for the
normal and rapid drawdown loading cases. The pseudostatic seismic stability
analyses indicate that the upstream slope has marginal stability for MDE seismic
loadings. This consideration, coupled with the possibility of liquefaction of the
foundation soils, necessitates future earthquake stability review of the dam.
• The existing erosion protection on the upstream face of the dam has deteriorated
over the past 50 years. Some slope dressing work is required to restore the rip rap
slope protection. A few areas in the downstream face also require some minor
slope dressing work.
• Only limited seepage could be observed in the right bank of the river and in the
left bank tunnels in the area downstream from the dam. This is evidence of the
effectiveness of the dam and its grout curtain. There is, however, an unexplained
hole in the el 980 bench roughly on line with an area where there was reported
seepage from the toe in the 1950s. This feature should be assessed in the field and
investigated further.
• It is not clear if the crest of the dam was raised during the 1979-80 construction
program. The current crest elevation and the profile of the top of the core are
uncertain.
• There are no operable means or instrumentation to measure settlement and/or
displacement within the dam or hydraulic gradients within the dam core.

Spillway
7-2
• The uncompleted spillway concrete structure is in fair to good condition and can
most probably be incorporated into the final structure.
• No records are currently available for the design of the spillway structure which
was started in the late 1970s. Design construction details must be found and
testing of the structural fabric (concrete and rebar) must be carried out before a
meaningful technical review of the spillway structure can be completed
• The excavated spillway channel rock slopes have suffered some deterioration over
the years. These are in a hazardous condition in some areas above future work
sites for the planned gate installations. Some remedial works, including scaling of
loose blocks and localized rock support are needed.
Miscellaneous
• The site area has experienced military activity in the past. Some de-mining work
has been carried out during the past year. There are uncleared mine areas near the
granular stockpile areas. Unexploded ordinance (UXO) was seen at many
locations near the dam, spillways and intake areas. In particular, there is an
unusually large quantity of UXO in the emergency spillway. UXO presents a
hazard to future site studies and remedial works throughout the project area.
7.2 Recommended Additional Investigations and Studies
The current dam condition investigation has identified a number of areas in which
more information is needed in order to complete a engineering assessment of the dam
and related structures. Recommended studies are described in the following
paragraphs. Table 7.1 lists the estimated time durations and specialists required for
each task.

Table 7.1 Investigation of the Kajakai Dam and Spillway
Activity
Work Duration (days)
Home Support
Staff
Geotechnical Seismicity Hydraulic Civil Surveyor
Secr. Draft Afghanistan Home Afghanistan Home Afghanistan Home Afghanistan Home Afghanistan Home
1) Surveys - - 2 - - - - - - - 8 4
2) 980 Berm - - 4 1 - - - - - - - -
3) Instrumentation
assessment
- - 3 2 - - - - - - - -
4) Monitoring
monuments and
construction benchmarks
- - - - - - - - - 3 -
5)Find and review
documents and
Drawings
- - 2 2 - - - 4 2 2 - -
6)Dam Crest
Investigation
- - 3 2 - - - - - - - -
7) Seismicity study - - - - - 25 - - - - - -
8) Seismic stability of
dam
- - - 8 - - - - - - - -
9) Hydraulic engineering
review
- - - - - - - 10 - - - -
10) Inspection of
concrete structure
2
11) Sampling of
concrete and steel of
spillway structure
- - - - - - - - 4 2 - -
12) Time in Kabul to
expedite work
- - 5 - - - - - 3 - 3 -
13) Mob/demob.
Afghanistan
- - 5 - - - - - 5 - 5 -
Report preparation 8 12 - 15 - 5 - 5 - 4 - 2
Total 8 12 24 30 - 30 - 19 16 8 19 6
1

Table 7.1 Assumptions:
1) Concrete core rig is rented to Canada for Item 10, concrete sampling. Portable generator is purchased or rented in Kabul by Louis Berger Group prior to
arrival of Acres staff. Both are transported to the site. Laborers will be provided by the powerhouse staff to assist Acres with this work.
2) Two test pits will be manually excavated in the crest of the dam across the centreline. This work is carried out by site laborers under supervision of
expatriate engineer. The purpose of the test pits is to provide samples of the core material.
3) Item 12, time in Kabul, covers expediting the field program, arranging for sample testing and waiting for transportation.
2

7-4
1) Surveys: Carry out a limited topographic survey in the dam and spillways area.
This survey will delineate the following:
a) Produce a detailed profile along the crest of the embankment dam.
b) Locate the seepage points in the cliff face downstream of the right abutment of
the abutment dam.
c) Survey the emergency spillway to delineate the outline of the channel in plan
and determine the topography of the uncompleted channel floor. The survey
should locate the piles of blasted rock in the spillway channel and note the
locations. The terrain downstream of the existing excavation should be
included in the survey. There had been earlier concerns of erosion of
unconsolidated materials deposited in this area.
d) Survey the concrete structure in the main spillway channel.
e) Survey the location of the fault zone in the spillway channel, just upstream
from the dam.
2) Investigation of el 980 berm in downstream face of the dam: Excavate a test
pit in the depression in the el 980 berm. This purpose of this test pit is to
investigate the extent and the cause of the cavity at this location. This test pit
must be carried out under the continuous direction of a qualified geotechnical
engineer, who will provide logs and sketches of the cavity as the excavation
progresses. It is important that the geotechnical engineer be present during this
work since the excavation of the test pit can destroy important details of the
cavity’s characteristics.
3) Dam Piezometer System Assessment: Determine what types of instruments and
leads have been installed. Determine which, if any of the instruments are
operable. Reestablish regular readings of working instruments if this is
considered feasible. If the instrumentation system cannot be salvaged several
piezometers should be installed in boreholes in the downstream side of the core in
order to monitor piezometric pressure changes when the dam is fully impounded
after completion of the spillway gates.
4) New Surface Monuments in the Dam and Construction Benchmarks:
a) Install a network of about 10 surface monuments along the crest and abutment
of the dam. Establish a program of regular surveys of these monuments to

7-5
provide background data prior to raising the reservoir. The monitoring should
continue after the reservoir has been raised.
b) Install at least two construction bench marks at each of the powerhouse and
spillway sites.
5) Documents and Drawings: Design documents and drawings pertaining to
the1978-80 spillway gates project should be obtained and reviewed. These are
essential if the spillway gates are to be completed as originally designed. With
this data, it will be possible to complete the current engineering assessment of the
project.
6) Test Pits in the Dam Crest: Earlier test pits were located too far downstream to
enable an evaluation of the top of the core. It is recommended that two test
trenches be excavated across the crest. These trenches should be 2.5 to 3.0 m
deep and be at least 5 m in length. They should straddle the centreline of the dam.
The work should be carried out under the supervision of an experienced engineer.
The test pit should not cut completely across the top of core. When backfilling the
test pit, the core material should be recompacted in 15 cm lifts with a mechanical
hand held tamper.
7) Seismicity Study: Site specific studies should be carried out to identify the
Maximum Credible Earthquake (MCE), probability of ground motion recurrence
curves, peak ground acceleration (PGA) and other criteria needed to evaluate the
seismic stability of the embankment dam. The study will include the following.
a) Assess the catalogue of instrumental earthquake event for the area within a
300 km radius of the damsite.
b) Literature review of the seismotectonics, historical earthquake history and
geology of the region.
c) Assess the activity and earthquake capability of local and regional faults.
d) Probabilistic analysis to establish ground motion recurrence relationships for
the damsite.
e) Deterministic study to establish the MCE and associated PGA for the site.
f) Establish maximum design ground motion parameters for the dam stability
analysis.

7-6
8) Seismic Stability of the Dam: Further seismic slope stability studies should be
carried out. This work will be an extension of the slope stability analyses already
carried out but will include the results of the seismicity studies as detailed in
recommended study No. 5 above. This work will included the following tasks:
a) Carry out a liquefaction potential review of the embankment dam and its
alluvial foundation. This review will include an assessment of the soil and
rockfill properties using old drilling information, geotechnical records from
the damsite and other sources. Liquefaction assessments will follow the
procedures outlined by Seed (7) and by the Southern California Earthquake
Centre (8).
b) Perform a pseudo-dynamic stability analysis to determine the range of
embankment deformations which will be caused by the design earthquake
ground motions. Assuming that serious liquefaction problems are not
identified, this will consist of a Newmark type sliding block analysis, which
will determine the displacement resulting from earthquake ground motions.
Procedures will follow those outlined by Makdesi and Seed (5). The
identification of serious liquefaction potential would necessitate a full
dynamic analysis of the embankment. It should be noted that the analysis
estimate in Table 7.1 assumes that the simplified sliding block procedure is
followed. Additional work would be required to perform a rigorous dynamic
analysis.
9) Hydraulic Engineering Review. Carry out a hydraulic assessment of the
following aspects of the project design.
a) Assess the requirement for the channel shaping excavation at the downstream
end of the natural spillway outlet channel where it intersects the Helmand
River, downstream from the powerhouse. This study will determine if there
will be hydraulic interference or erosion in the tailrace area when the spillway
gates are installed and the reservoir is operated at its original design levels.
b) Review the criteria for the emergency spillway and fuse plug design, once the
1978-80 designs are available. The work estimates given on Table 7.1 assume
that the original design data are available.
10) Spillway Structure Materials Testing: A limited program of materials testing
should be included in the final engineering assessment of the spillway structure.
This will consist of:

7-7
a) Obtain concrete core samples from representative areas of the concrete
structure and carry out a total of 10 unconfined compressive tests. The
sampling and testing will follow the procedures outlined in ASTM C42.
b) Carry out 10 tensile strength tests on samples of rebar from the structure. The
tests should be carried out in accordance with the procedures outlined in
ASTM A 370.
11) Reservoir Rim Study: Carry out a reconnaissance to establish the reservoir rim
stability for the new impoundment levels. The reconnaissance will examine
terrain and geological aspects of the rim which determine the susceptibility of the
slopes for instability. The study will also document the few dwellings and areas
of agriculture which may be located on areas, which will be subject to new
flooding. The reconnaissance will be carried out by means of a boat or helicopter
survey.
12) Mines and Unexploded Ordinance (UXO). Appropriate action must be taken by
qualified personnel to remove all explosive hazards form the site as soon as
possible.
7.3 Recommended Remedial Work
A few minor remedial items have been identified. This work is not urgent and can be
deferred until after a more complete engineering evaluation has been carried out.
1 Riprap and Slope Dressing: Slope dressing work is required to remediate
deterioration of the upstream berms and inter-berm faces as well to carry out
localized treatment of the downstream slopes. The work will clear boulders from
the berm surfaces and rearrange the riprap slope protection of the slopes. The
slope dressing will be carried out with a back hoe. The objective of the work will
be to redistribute the rip rap material into a relatively uniformly graded layer of
interlocked rock fragments. This work will use the rock fill material that is
presently on the slope and no new material will need to be quarried.
2 Scaling and Slope Stabilization in the Spillway: Scaling work is needed in the
slopes on both sides of the spillway. The exact extent of the work will be
determined when the final design drawings of the spillway gate structure are
available. It is probable that a few rock bolts will be needed to stabilize localized
rock features.

References

References
1) Blueifuss, D.J., Haeke, J.P., Rockfill Dams: “Design and Construction
Problems”, Paper No. 3072, Transactions of the American Society of Civil
Engineers, 1954, (illustrated with discussions on design and construction of
Kajakai dam).
2) Federal Energy Regulatory Commission (FERC) “Engineering Guidelines for
the Evaluation of Hydropower Sites”, April 1991 (updated in November
2003).
3) Harza Overseas Engineering Company, “Contract Documents 912-2, Volume
III, Technical Specifications”, April, 1977.
4) ICOLD, “Selecting Seismic Parameters for Large Dams”, Bulletin 72, 1988.
5) International Engineering Company Inc. “ Feasibility Study for Installation of
Spillway Gates at Kajakai Reservoir, Afghanistan”, August, 1971.
6) Makdisi, Seed, Simplified Procedure for Estimating Dam and Embankment
Earthquake-Induced Deformations, Journal of Geotechnical engineering
division, ASCE, Vol. 104, No. GT7, 1978.
7) Morrison-Knudsen Afghanistan, Inc., “Final Design report on Kajakai Dam,
Arghandab Dam and Boghra Canal Projects”, prepared for Morrison-Knudson
Afghanistan Inc. by International Engineering Company Inc., December,
1956.
8) Seed, H.B., and Idriss, I.M., “Simplified Procedure for Evaluating Soil
Liquefaction Potential,” Journal of the Soil Mechanics and Foundations
Division, ASCE, Vol 97, No. SM9, 1971.
9) Southern California Earthquake Centre, University of Southern California,
Recommended Procedures of Implementation of DMG Special Publication
Guidelines for Analyzing and Mitigating Liquefaction in California. 1999.

Appendix A
Photographs

Appendix A
Spillway Area Photographs
February 10 to 14, 2004
Photo S1: Aerial view of Kajakai spillway looking upstream. Photo shows the
uncompleted gate foundation/rollway structure and the downstream stilling basin.
Note the cut-out for the fault excavation works in the upstream apron slab.
Photo S2: Aerial view of spillway looking downstream

A-2
Photo S3: Top of uncompleted concrete rollway structure.
Photo S4: Uncompleted rollway, Bay 5

A-3
Photo S5: Profile of partially completed rollway. Waterstop is visible in
upper edge of rollway.
Photo S6: Typical condition of exposed rebar.

A-4
Photo S7: Downstream side of stilling basin
Photo S8: Right bank of the spillway excavation. The J1 joints are visible in
the rock face. Note the relatively unjointed rock mass at the end of the
overflow deck.

A-5
Photo S9: Karstic J1 joints in the right abutment of the spillway, located
immediately downstream of the concrete gate foundation structure. These joints,
which are infilled with 1 to 30 cm of clay, are aligned oblique to the gate structure
axis and cross under the base of the foundation.
Photo S10: Karstic J1 joint in the left side of the spillway, immediately
downstream of the concrete structure.

A-6
Photo S11: Spillway left bank. Note the karstic J1 joints in the rock slope,
located upstream and downstream of the deck of the concrete structure.
Photo S12: View looking at the spillway right bank and the downstream concrete
lining of the stilling basin.

A-7
Photo S13: Right bank and base of the spillway stilling basin. Note the rock debris
which partially covers the base of the stilling basin. Portal to right bank gallery is
visible just right of the centre of the photograph.
Photo S14: Portal of the right bank stilling basin gallery.

A-8
Photo S15: Karstic fault zone located on the upstream edge of the foundation
structure in the centre of the spillway channel
Photo S16: View looking upstream at the upstream karstic fault zone. Note the
cutout in the upstream apron.

A-9
Photo S17: Karstic cavity in the left side of the fault zone, just upstream of
the upstream edge of the concrete foundation structure.
Photo S18: View of karstic cavity and the solutioned subvertical J1 joint.
Note upstream edge of concrete backfill bulkhead in the lower right side of the
photograph, just above the notebook.

A-10
Photo S19: View looking upstream from downstream end of the spillway
channel. Note the karsified gully at the downstream end of the excavated
channel.
Photo S20: Panoramic view of the partially excavated emergency spillway.
Note the abandoned excavation equipment.

A-11
Photo S21: Stockpiles of granular material, located northwest of the main
spillway channel.

Kajakai Reservoir Photographs February 10 to 14, 2004
Photo R1: Rocky Narrows. This constriction divides the reservoir into upstream and
downstream sections. View looking downstream towards the Downstream Reservoir. Note
the cloudy water in the foreground Upstream Reservoir and along the right side of the
narrows channel. Located approximately 15 km upstream from the Kajakai Dam
Figure R2: Floating algae in the Rocky Narrows. Note the relatively cloudy water in the
lower right side of the photo.

Kajakai Reservoir Photographs February 10 to 14, 2004
Photo R3: Upstream Reservoir, looking upstream. The rocky narrows are on the left side of the photograph. The large alluvial fan makes up
most of the eastern shoreline in the right side of the photo. Rugged rock uplands rise above the western and northern shorelines. Note the
cloudy plume in the water along the northwest (far) side of the reservoir. This cloudy water extends through the Rocky Narrows on the left side
of the photograph.

Kajakai Reservoir Photographs February 10 to 14, 2004
Photo R4: Upstream Reservoir, southeast shore. Small village and cultivated land adjacent located on the
alluvial fan deposits.
Photo R5: Upstream Reservoir, looking downstream towards the southwest. Prominent terrace scarp is visible
in the foreground of the photo. Small scale, local slope slumps can be expected in this scarp when the reservoir
level fluctuates. Rocky Narrows in the background

Kajakai Reservoir Photographs February 10 to 14, 2004
Photo R6: Downstream Reservoir, view from the north end looking downstream towards the west. Note the rugged rocky terrain on both sides of the reservoir.
Photo R7: Downstream Reservoir. View looking north towards the north shoreline.

Kajakai Reservoir Photographs February 10 to 14, 2004
Photo R8: Downstream Reservoir. View looking southwest showing the south shoreline on the left side of the photograph. Damsite is located in the rock ridge forming the
southwest shoreline on the right side of the photograph. Clearly visible is the low relief terrain of soft shaly rock which makes up most of the south shoreline of the
Downstream Reservoir. Rugged limestone bedrock terrain consisting is visible on the right side of the photo.
Photo R 9: Small village with cultivated land on the northwest shore of the Downstream Reservoir, located about 5.5 km from of the Kajakai Dam.

Appendix B
Seismicity Review

Table of Contents
B1 INTRODUCTION............................................................................................................. B-1
B2 AFGHANISTAN SEISMICITY ...................................................................................... B-2
B3 HISTORICAL EARTHQUAKE DATA ......................................................................... B-3
B3.1 1973 TO 2004 NEIC DATA......................................................................................... B-3
B3.2 AMBRASEYS’ EARTHQUAKE RECORDS - PRIOR TO 1975 ............................................ B-4
B4 ASSESSMENT .................................................................................................................. B-4
B5 DESIGN EARTHQUAKE PARAMETERS................................................................... B-5
B5.1 OPERATING BASIS EARTHQUAKE............................................................................... B-5
B5.2 MAXIMUM DESIGN EARTHQUAKE.............................................................................. B-5
B6 SEISMIC COEFFICIENT FOR PSEUDOSTATIC STABILITY ANALYSES ......... B-8
B7 CONCLUSIONS ............................................................................................................... B-9
Tables

Appendix B
Seismicity Review
B1 Introduction
A preliminary seismicity review has been carried out as part of the Kajakai Dam
condition assessment. The purpose of this study was to assess the general level of
seismicity in the Kajakai area and to establish a realistic maximum design
earthquake for use in stability analyses. The study used available historical
records and technical data from various sources. No previous earthquake study
was carried out for the damsite area and published seismicity information for
southern Afghanistan is scarce.
The study included the following tasks.
• Establish the basic seismic zoning of the region and estimate 1/475 Peak
Ground Acceleration (PGA) using the seismic hazard map of the Global
Seismic Hazard Assessment Program (GSHAP).
• Collect earthquake data for all earthquakes within a 300 km radius of the
Kajakai site.
• Deterministically establish maximum earthquake events which affect the
Kajakai site.
• Establish the Operating Basis Earthquake (OBE) based on the GSHAP
seismic hazard maps.
• Deterministically establish the Maximum Design Earthquake (MDE) using the
seismic zonation and earthquake records.
• Calculate the PGA and pseudostatic seismic coefficients associated with the
MDE.
Realistic design parameters have been obtained from the current seismicity study
work. These criteria have been used for the stability analyses of the embankment

B-2
dam. It must be emphasized that this is a preliminary study, which relied on a
deterministic assessment of simplified seismicity models. A more rigorous
assessment must be carried out at a future date.
The results of the seismicity study are presented in the following paragraphs and
on the attached tables and figures.
B2 Afghanistan Seismicity
Afghanistan, which lies on the southern fringe of the Eurasian plate, borders the
Arabian plate to the south and the Indian plate to the southeast. Both of these
plates are moving at rates of 20 mm/yr to 40 mm year relative to the central
Afghanistan plate. These regional movements and related tectonic activities
trigger large numbers of earthquakes within the country and in the surrounding
regions.
Earthquakes occur with an uneven distribution across Afghanistan. The central
areas of the country are generally seismically inactive. Large areas in the
northeast southeast and west parts of the country have high rates of seismic
activity. Very high rates of seismicity with repeated destructive earthquakes are
characteristic of
• northeast areas within and near the Hindu Kush, including the regions around
Kabul,
• the east and south east areas, including Pakistani Baluchistan, and
• the western borders with Iran.
Figure B1 presents the seismic hazard map of Afghanistan, which is based upon a
compilation made by the GSHAP. This figure shows that the Kajaki site is within
the relatively low seismic central region of the country. This area has been
classified as a zone of “moderate seismic hazard”. Local earthquakes are rare in
this region. Most seismic hazard is the result of distant earthquakes from the
southeast near Pakistan and from the western border area near Iran.

B-3
B3 Historical Earthquake Data
Figure B2 shows the distribution of earthquakes which have occurred with a 300
km radius of the site. This data has been obtained from two earthquake
catalogues.
• Geological survey of America NEIC data base: 1973 to 2004
• Ambraseys: Historical data prior to 1973.
B3.1 1973 to 2004 NEIC Data
Earthquake data has been obtained from the NEIC US Geological Survey
Earthquake Database for earthquakes within a 300 km radius of the site. This
record, which covers the period from 1973 to 2004, is given on Table 3.1. Figure
B2 shows the distribution of the earthquake epicenters.
During the 30 years of record, a total of 75 earthquakes occurred within a radius
of 300 km from the site. The majority of these events were clustered in a
northeast trending belt, located about 200 to 300 km east and southeast from the
Kajaki site. These earthquake epicenters are clustered within the High Hazard
zone shown on Figure B2. In addition to events in this zone, a few randomly
distributed events occurred about 150 to 200 km north, south and west of the
Kajaki site. Only two of the earthquakes occurred within 150 km of the Kajakai
damsite.
The magnitudes of the 75 NEIC earthquakes are distributed as shown in the
following table.
Magnitude No. Events
Unknown 10
< 4 14
4 to 5 43
5 to 6 6
> 6 2
The largest earthquake in the NEIC record was a M 6.7 earthquake, which was
located 249 km southeast from Kajaki. This earthquake occurred on October,
1975. The closest significant event was an M 4.6 earthquake, which occurred
130 km approximately to the west of the dam in 2001.

B-4
B3.2 Ambraseys’ Earthquake Records - Prior to 1975
Table 3.2 lists all historical and significant earthquakes located in the region
within at east latitude 30 to 34 deg and north longitude 62 to 68 degrees as listed
by Ambraseys (2003) for the period up to 1973. These events are plotted together
with the modern NEIC 1973 to 2004 events on Figure B2. The original
Ambraseys listing includes all known historical earthquakes for Afghanistan.
Only 11 Afghanistan events listed by Ambraseys are located within the study
area.
• The geographical distribution of historical earthquake epicentres closely fits
the patterns of the instrumental events recorded in 1973 to 2004.
• Ambraseys et al have assigned magnitude values to the better documented
events. All of these earthquakes are larger than M 5.
• An event of unknown magnitude occurred near Kandahar in 1857. This event,
which was located approximately 80 km southeast of Kajaki, is the closest
earthquake to the damsite.
• In 1933 a M 5.7 earthquake occurred in Uruzgan. The epicentre of this
earthquake was approximately 200 km east from Kajaki.
B4 Assessment
A simplified model has been assumed. This model considers only the seismicity
of the broad areas and has not assessed the earthquake characteristics of
individual tectonic elements such as faults. Further, more detailed work will
undoubtedly assess more source zones, but the current model is deemed sufficient
for the current preliminary study.
The available data indicates that the Kajaki site is influenced by earthquakes from
two sources:
• Central “Moderate Hazard Zone”: Earthquakes are randomly distributed
throughout this area. There is not enough data to determine if the earthquakes
in this zone are related to any particular geological features. It is assumed that

B-5
small to medium sized “floating” earthquakes of M 4 to M 5.5 can occur
anywhere within the central block, including in the immediate vicinity of the
damsite.
• Southeast “High Hazard Zone”: Over 90 percent of the earthquakes in the
study area have occurred in a northwest trending zone, which extends from
about 150 km to more than 300 km east and southeast of the Kajaki site. This
area covers a zone which is distributed on both sides of the Afghanistan/
Pakistan border. This area is characterized by high frequency seismicity and
has hosted numerous earthquakes of M 7.0 and larger.
B5 Design Earthquake Parameters
In view of the preliminary nature of the seismic studies, dam stability was
checked for a range of seismic coefficients. Two levels of earthquake ground
motion are used for the design dams, abutments and associated works (ICOLD
Bulletin 72).
• OBE
• MDE.
B5.1 Operating Basis Earthquake
The OBE design is used to limit the earthquake damage to a dam project. This
event traditionally has a return period of about 145 to 500 years. ICOLD
specifies that the dam shall remain operable after the OBE and only minor easily
repairable damage is accepted. The OBE used in this study is adapted from the
Global seismic Hazard Assessment Program (GSHAP) data shown on Figure B1.
This map shows the ranges of PGA values with for a return period of 475 year
return period. The OBE is appropriate for buildings and most civil structures
which are normally designed to building code seismic criteria.
B5.2 Maximum Design Earthquake
The dam must be stable for the MDE. ICOLD Bulletin 72 specifies:

B-6
“For dams whose failure would present a great social hazard the MDE
will be characterized by a level of motion equal to that expected at the
dam site from a deterministically evaluated MCE (Maximum Credible
Earthquake) or of the earthquake determined using probabilistic
procedures…”
Ideally both probabilistic and deterministic methods are used to determine the
MDE for a significant dam like Kajaki. Given the preliminary nature of this
study, the MDE has been selected by means of a simplified deterministic
assessment of the MCE. More detailed seismicity studies should be carried out
during the second phase dam safety assessment.
ICOLD and other design agencies state that that the MDE shall not cause any
uncontrolled release of water from the reservoir. Significant structural damage is
accepted from an MDE event.
The MCE, which is the basis of the MDE, is the largest reasonably conceivable
earthquake that appears possible along a recognized fault or within a
geographically defined tectonic province, under the presently known or presumed
tectonic framework.
The following source areas are considered when estimating the MCE for the
Kajaki site.
1) Local Background Seismicity: The site can be affected by a local floating
event. The largest instrumented earthquake that occurred in the “moderate
hazard” area, within 150 km of the damsite is an M 4.6 event. This
earthquake occurred in 2001 and was located 130 km west of Kajaki. A few
smaller events of unknown, but comparable, magnitude have occurred within
about 100 km of the site. Based on the currently available data, it is assumed
that that the largest credible event for the moderate hazard zone is in the range
of M 5.0 to M 6.0. Given the absence of known earthquakes within about 80
km of the site, it is assumed that the site may be subjected to the effects of a
near field M 5.5 earthquake.
2) Southeast High Hazard Zone: The high hazard zone is located southeast of
Kajaki and lies between Kandahar and the Quetta area in Pakistan. The data
on Table 3.1 show that this zone has experienced an M 6.7 event at a distance

B-7
of 249 km from Kajaki during the past 30 years. As seen on Figure B2, the
area of high frequency seismic activity lies as close as about 150 km to the
Kajakai site. Earthquake records for the past 100 years indicate that numerous
events of M >7.0 have occurred within this high hazard zone inside Pakistan,
(further than 300 km from the site). This zone is considered to have a
maximum earthquake of M 8.0.
This preliminary assessment indicates that two MCE events should be considered
when calculating the MDE for design.
• Near field M 5.5 earthquake, which has a hypocentral distance of 15 km from
the damsite. This is considered to be the maximum earthquake for the
“moderate risk” area surrounding the damsite.
• Far field M 8.0 event, which occurs 150 km from the damsite. It is assumed
that this earthquake is generated in the high hazard zone located east and
southeast of Kandahar.
Suitable attenuation relationship must be used to determine the PGA at the site.
Two attenuation formulas are used herein to calculate PGA ground motions.
• Joyner and Boore (1981), suitable for near field events
• Youngs (1988), far field subduction events.
The following preliminary PGA values have been calculated.
Earthquake Peak Ground Acceleration (g)
Magnitude Hypocentral Distance Joyner and Boore Youngs
8.0 150 Na 0.17
5.0 15 0.25 na
As can be seen, the near field M 5.5 earthquake produces a larger PGA than the
far field M 8.0 event. The larger PGA value of the near field earthquake should
be used for deriving the seismic coefficient for pseudostatic analysis. It should be
noted, however, that the more distant M 8.0 earthquake also has a significant
destructive potential because of its longer time duration and long period ground
vibrations. Both earthquakes should be considered when carrying out more
sophisticated dynamic analyses.

B-8
It is emphasized that the MCE events elected herein are preliminary criteria,
which are subject to revision when more detailed seismicity studies have been
carried out.
B6 Seismic Coefficient for Pseudostatic
Stability Analyses
The PGA value represents the highest spike of ground motion on an earthquake
acceleration record. This value is too high to use in pseudostatic stability
analyses, which assume that the seismic loading is constant. It is customary uses
a sustained seismic coefficient which is less than the PGA value. It is
conservatively assumed herein that the pseudostatic seismic coefficient (k) is
equal to two thirds of the PGA.
The following PGA and seismic coefficient values are recommended for the
Kajakai site.
Earthquake
Operating Basis
Earthquake (OBE)
(1/ 475 year event)
Maximum Design
Earthquake (MDE)
Peak Ground
Acceleration
(PGA, g)
Seismic coefficient, k,
for Pseudostatic
Analyses
(2/3 x PGA)
0.16 0.11
0.25 0.17
The OBE gives a stability check for lower levels of expected seismic events and is
compatible with building code criteria. It should be used for the analysis of civil
structures and other facilities whose failure would not lead to an uncontrolled loss
of the reservoir.
The MDE earthquake permits the most conservative estimate of the effects of
earthquake loadings. The PGA value of 0.25 g for the MDE event is equal to
greater than the maximum ground motion caused by either a near field M 5.5 near
field earthquake or a far field M 8.0 event. The return period of this PGA is not
known, but is expected to be in the range of 2000 to 8000 years. A site specific

B-9
probabilistic analysis should be carried out at some future date in order to
determine the return period of the MCE.
B7 Conclusions
The Kajakai site lies within an area of moderate seismic hazard. Very few
earthquakes have been recorded within 150 km of the site. There is a high hazard
zone of fairly intense seismic activity located from about 150 km to more than
300 km southeast of Kajakai. Most earthquakes in Southern Afghanistan occur in
this zone at distances of 200 to more than 300 km from the site.
Two sources of earthquakes contribute to the seismic hazard at the Kajakai site.
• Near field earthquakes of up to M 5.5 are assumed to potentially occur as
close as 15 km from the site.
• Far field M 8.0 earthquakes which can occur as near as 150 km from the site.
The current preliminary studies have identified two levels of earthquake for the
analysis of structures at the Kajakai site.
• OBE, characterized by a PGA of 0.16 g. A pseudostatic coefficient (k) of
0.11 g is recommended for this 1/475 year event. The PGA for this event was
taken from the GSHAP hazard map. The OBE should be used for analysis of
civil structures and other facilities normally designed with “building code”
criteria.
• MDE, characterized by a PGA of 0.25 g. A pseudostatic coefficient (k) of
0.17 g is recommended for this deterministic maximum credible earthquake
event. The MDE should be used for the analysis of the dam and related water
retention facilities whose failure would lead to an uncontrolled loss of the
reservoir.

Tables

Table B1
NEIC: Earthquake Search Results
U. S. Geological Survey
Earthquake Data Base
FILE CREATED: Wed Feb 25 08:12:18 2004
Circle Search Earthquakes= 75
Circle Center Point Latitude: 32.200N Longitude: 65.000E
Radius: 300.000 km
Catalog Used: PDE
Data Selection: Historical & Preliminary Data
CAT
YEAR
MO
DAY
ORIG
TIME
LAT
LONG
DEPTH
MAG
UDE
DIST
km
PDE 1973 2 4 192548.3 31.67 67.26 40 3.9 GS 221
PDE 1975 7 18 93012.2 30.97 66.70 35 4.9 GS 210
PDE 1975 9 28 62146.9 30.21 66.38 35 256
PDE 1975 9 28 70838.9 30.14 66.19 6 254
PDE 1975 9 28 101940.3 30.26 66.29 8 4.2 GS 247
PDE 1975 10 3 51423.3 30.25 66.32 11 6.7 GS 249
PDE 1975 10 3 173135.8 30.41 66.35 33 6.4 GS 235
PDE 1975 10 3 175503.7 30.42 66.47 33 4.9 GS 241
PDE 1975 10 3 190807.3 30.47 66.49 23 4.9 GS 238
PDE 1975 10 4 55942.9 30.33 66.23 15 4.5 GS 238
PDE 1975 10 6 125137.7 30.13 66.26 24 4.4 GS 258
PDE 1975 11 4 170943.3 30.13 66.39 31 264
PDE 1975 11 7 54128.4 30.88 66.60 40 210
PDE 1976 2 26 182541.8 30.56 66.48 33 229
PDE 1976 8 20 112202.2 30.45 66.38 33 4 GS 234
PDE 1977 11 18 63902.1 30.25 66.27 27 4.9 GS 247
PDE 1978 2 23 160237.1 31.50 67.01 33 4.6 GS 205
PDE 1978 3 16 20000.5 29.93 66.30 33 5.9 GS 280
PDE 1978 4 19 101923.2 29.84 66.27 26 4.2 GS 288
PDE 1978 5 6 111609.9 29.84 66.21 33 5.7 GS 285
PDE 1978 12 4 1850.9 30.26 66.34 33 4.3 GS 250
PDE 1979 1 31 155033.9 29.90 63.87 183 4.8 GS 276
PDE 1979 6 24 134903.4 30.30 66.35 40 4.8 GS 246
PDE 1979 11 27 123933.4 32.67 67.15 33 4.8 GS 208
PDE 1980 7 3 223005.1 30.87 67.68 33 4.3 GS 293
PDE 1980 10 29 195542.2 31.93 67.04 33 3.9 GS 194
PDE 1981 5 7 62859.8 31.35 66.69 33 3.8 GS 185
PDE 1984 4 11 55758.01 32.17 67.05 33 4.3 GS 193
PDE 1984 4 11 123429.81 30.51 66.41 33 3.9 GS 230
PDE 1985 3 19 91024.37 30.06 66.32 33 4.2 GS 268
PDE 1985 4 15 112400.45 30.25 66.33 33 4.8 GS 250
PDE 1985 11 4 123723.71 31.80 67.54 33 4.5 GS 243
PDE 1986 1 22 20116.26 30.20 66.31 33 4.4 GS 253
PDE 1986 2 11 165802.48 30.04 66.40 33 4.1 GS 274

Table B1
NEIC: Earthquake Search Results
U.S. Geological Survey
Earthquake Data Base - 2
CAT
YEAR
MO
DAY
ORIG
TIME
LAT
LONG
DEPTH
MAG
UDE
DIST
km
PDE 1986 2 19 43958.48 32.11 67.66 33 4.3 GS 251
PDE 1986 2 19 82040.4 33.13 67.17 33 4.3 GS 227
PDE 1987 8 10 105219.94 29.87 63.84 164 5.6 GS 281
PDE 1989 1 26 182229.51 31.75 67.18 33 4.2 GS 211
PDE 1989 3 28 204038.55 33.74 65.11 23 4.6 GS 171
PDE 1989 3 31 33312.07 31.27 66.87 10 4.2 GS 205
PDE 1989 4 11 52300.18 29.72 64.09 33 4.6 GS 287
PDE 1990 6 15 55455.74 30.58 67.43 33 4.4 GS 292
PDE 1990 11 6 233823.53 34.13 63.22 179 4.5 GS 270
PDE 1991 2 25 23842.85 30.84 67.20 33 4.7 GS 257
PDE 1991 2 25 200139.25 30.62 67.50 33 4.5 GS 295
PDE 1991 2 26 11927.39 31.35 67.20 33 4.3 GS 228
PDE 1991 9 15 2050.3 30.62 66.74 33 4.8 GS 240
PDE 1991 9 15 21224.94 30.72 66.76 25 4.6 GS 233
PDE 1992 1 6 155906.95 30.67 65.99 33 193
PDE 1992 2 5 193629.88 31.51 67.04 33 4.4 GS 207
PDE 1992 2 5 231048.63 31.43 66.82 17 5.3 GS 192
PDE 1992 2 5 234136.85 31.36 66.86 33 5 GS 198
PDE 1992 7 14 42428.28 30.18 66.32 33 4.6 GS 256
PDE 1992 11 17 23850.1 33.78 67.57 33 5.2 GS 297
PDE 1993 3 11 425.51 31.53 67.10 13 4.7 GS 211
PDE 1993 7 26 93130.34 29.96 66.63 33 4.9 GS 292
PDE 1993 11 16 155248.52 30.80 67.22 26 5.6 GS 261
PDE 1996 12 16 50327.12 33.26 64.51 33 126
PDE 1997 1 24 234004.43 33.04 67.85 33 283
PDE 1997 6 20 230443.95 29.64 65.79 33 4.1 GS 293
PDE 1997 7 8 192658.52 31.29 66.63 33 184
PDE 1997 8 25 141322.49 31.06 67.77 33 3.8 GS 291
PDE 1998 1 13 144019.29 32.56 67.05 33 196
PDE 1998 3 3 1144.82 31.16 67.56 33 3.7 GS 268
PDE 1998 5 2 34929.92 31.06 67.77 33 3.8 GS 291
PDE 1999 2 23 40654.71 30.53 66.32 10 4 GS 223
PDE 1999 4 16 161437.21 30.81 67.49 33 3.8 GS 281
PDE 2000 1 22 91450.01 31.99 67.24 33 4.1 GS 212
PDE 2000 4 11 104659.99 31.02 67.74 33 3.8 GS 290
PDE 2000 8 22 1431.49 30.60 66.65 33 4.2 GS 236
PDE 2001 2 22 75439.67 31.88 67.28 33 4.5 GS 218
PDE 2001 5 26 231438.05 31.74 63.73 33 4.6 GS 130
PDE 2001 8 14 211928.88 29.94 65.27 33 4.8 GS 251
PDE
PDE-
2002 3 16 200518.31 29.86 66.32 33 3.7 GS 288
W 2003 12 1 185328.62 31.79 67.89 33 3.7 GS 276

Table B2
Historical Earthquakes in the Kajakai Region
Historical Earthquakes for Latitude 30 to 34 deg and Longitude 62 to 68 deg
Y M D OT N E Ms h

Figures

40
35
30
60 65 70 75
0 100 200
Kilometers
60 65 70 75
0 0.2 0.4 0.8 1.6 2.4 3.2 4.0 4.8
Low
Hazard
Source:
Global Seismic Hazard Assessment Program (GSHAP)
Siesmicity of Afghanistan and Pakistan, 2003
Kajakai Dam
Kandahar
Moderate
Hazard
High
Hazard
Peak Ground Acceleration (m/s2)
10% Probability of Exceedance in 50 years, 475 year Return Period
Kabul
Very High
Hazard
Figure B1
Kajakai Project Seismic Review
Seismic Hazard of Afghanistan
40
35
30

40
35
30
60 65 70 75
Kajakai Dam
200km
Kandahar
Kabul
60 65 70
NEIC Data
Unknown M
M 3.5 to 4.0
M 4 to 5
M 5 to 6
M > 6
75
0 100 200
Kilometers
300km
100km
Data Sources:
(1) U.S. Geological Survey, NEIC Earthquake Database,
Historical and preliminary Earthquake data, data within
a 300km radius of Lat 32.2 deg, Long. 65.0 deg, for period 1975 to 2004
(2) Ambraseys, N and Bilhom, R., “Earthquakes in Afghanistan,
Research Letters in the Press, 2003, historical earthquakes for period 1850 to 1975
Ambraseys et. al. Data
Historical M>5
Figure B2
Kajakai Project Seismic Review
Earthquake Epicentre Locations
40
35
30

Appendix C
Dam Stability Analyses – Analysis Criteria
Slide Analysis Information

Appendix C
Dam Stability Analyses - Analysis Criteria
Slide Analysis Information
Project Settings
Project Title: SLIDE - An Interactive Slope Stability Program
Failure Direction: Right to Left
Units of Measurement: SI Units
Pore Fluid Unit Weight: 9.81 kN/m 3
Groundwater Method: Water Surfaces
Data Output: Standard
Calculate Excess Pore Pressure: Off
Allow Ru with Water Surfaces or Grids: Off
Random Number Generation Method: Park and Miller v.3
Analysis Methods
Analysis Methods used:
Bishop simplified
GLE/Morgenstern-Price with interslice force function: Half Sine
Number of slices: 25
Tolerance: 0.005
Maximum number of iterations: 50
Surface Options
Surface Type: Circular
Radius increment: 10
Minimum Elevation: Not Defined
Composite Surfaces: Enabled
Reverse Curvature: Create Tension Crack

Material Properties
Material: Sluiced Rock fill
Strength Type: Mohr-Coulomb
Unit Weight: 18.5 kN/m 3
Cohesion: 0 kPa
Friction Angle: 40 degrees
Water Surface: Water Table
Custom Hu value: 1
Material: Impervious Rolled Fill
Strength Type: Mohr-Coulomb
Unit Weight: 19.5 kN/m 3
Cohesion: 0 kPa
Friction Angle: 30 degrees
Water Surface: Water Table
Custom Hu value: 1
Material: Transition Zone
Strength Type: Mohr-Coulomb
Unit Weight: 19.5 kN/m 3
Cohesion: 0 kPa
Friction Angle: 35 degrees
Water Surface: Water Table
Custom Hu value: 1
Material: In-situ Alluvium
Strength Type: Mohr-Coulomb
Unit Weight: 21 kN/m 3
Cohesion: 0 kPa
Friction Angle: 38 degrees
Water Surface: Water Table
Custom Hu value: 1
C-2

C-3
Material: Compacted Free Draining Gravel
Strength Type: Mohr-Coulomb
Unit Weight: 19 kN/m 3
Cohesion: 0 kPa
Friction Angle: 40 degrees
Water Surface: Water Table
Custom Hu value: 1
Material: Limestone Bedrock
Strength Type: Mohr-Coulomb
Unit Weight: 27 kN/m 3
Cohesion: 100 kPa
Friction Angle: 55 degrees
Water Surface: Water Table
Custom Hu value: 1

C-4
Figure C1: Materials zoning of dam and foundation.
Figure C2: Upstream slope, steady state, normal loading.

C-5
Figure C3: Upstream slope, 0.11g seismic analysis.
Figure C4: Upstream slope, 0.17g seismic analysis.

C-6
Figure C5: Upstream slope, Rapid drawdown analysis.
Figure C6: Downstream slope, Normal loading analysis.

C-7
Figure C7: Downstream slope, 0.11 g seismic analysis.
Figure C8: Downstream slope, 0.17 g seismic analysis.