Volume 6 – Geotechnical Manual, Site Investigation and Engineering ...

GOVERNMENT OF MALAYSIA 

DEPARTMENT OF IRRIGATION 

AND DRAINAGE 

Volume 6 – Geotechnical 

Manual, Site Investigation and 

Engineering Survey 

Jabatan Pengairan dan Saliran Malaysia 

Jalan Sultan Salahuddin 

50626 KUALA LUMPUR

DID MANUAL Volume 6 

Disclaimer 

Every effort and care has been taken in selecting methods and recommendations that are 

appropriate to Malaysian conditions. Notwithstanding these efforts, no warranty or guarantee, 

express, implied or statutory is made as to the accuracy, reliability, suitability or results of the 

methods or recommendations. 

The use of this Manual requires professional interpretation and judgment. Appropriate design 

procedures and assessment must be applied, to suit the particular circumstances under 

consideration. 

The government shall have no liability or responsibility to the user or any other person or entity with 

respect to any liability, loss or damage caused or alleged to be caused, directly or indirectly, by the 

adoption and use of the methods and recommendations of this Manual, including but not limited to, 

any interruption of service, loss of business or anticipatory profits, or consequential damages 

resulting from the use of this Manual. 

March 2009 

i


Foreword 

The first edition of the Manual was published in 1960 and was actually based on the 

experiences and knowledge of DID engineers in planning, design, construction, operations and 

maintenance of large volume water management systems for irrigation, drainage, floods and river 

conservancy. The manual became invaluable references for both practising as well as officers newly 

posted to an unfamiliar engineering environment. 

Over these years the role and experience of the DID has expanded beyond an agriculturebased 

environment to cover urbanisation needs but the principle role of being the country’s leading 

expert in large volume water management remains. The challenges are also wider covering issues of 

environment and its sustainability. Recognising this, the Department decided that it is timely for the 

DID Manual be reviewed and updated. Continuing the spirit of our predecessors, this Manual is not 

only about the fundamentals of related engineering knowledge but also based on the concept of 

sharing experience and knowledge of practising engineers. This new version now includes the latest 

standards and practices, technologies, best engineering practices that are applicable and useful for 

the country. 

This Manual consists of eleven separate volumes covering Flood Management; River 

Management; Coastal Management; Hydrology and Water Resources; Irrigation and Agricultural 

Drainage; Geotechnical, Site Investigation and Engineering Survey; Engineering Modelling; 

Mechanical and Electrical Services; Dam Safety, Inspections and Monitoring; Contract Administration; 

and Construction Management. Within each Volume is a wide range of related topics including topics 

on future concerns that should put on record our care for the future generations. 

This DID Manual is developed through contributions from nearly 200 professionals from the 

Government as well as private sectors who are very experienced and experts in their respective 

fields. It has not been an easy exercise and the success in publishing this is the results of hard work 

and tenacity of all those involved. The Manual has been written to serve as a source of information 

and to provide guidance and reference pertaining to the latest information, knowledge and best 

practices for DID engineers and personnel. The Manual would enable new DID engineers and 

personnel to have a jump-start in carrying out their duties. This is one of the many initiatives 

undertaken by DID to improve its delivery system and to achieve the mission of the Department in 

providing an efficient and effective service. This Manual will also be useful reference for non-DID 

Engineers, other non-engineering professionals, Contractors, Consultants, the Academia, Developers 

and students involved and interested in water-related development and management. Just as it was 

before, this DID Manual is, in a way, a record of the history of engineering knowledge and 

development in the water and water resources engineering applications in Malaysia. 

There are just too many to name and congratulate individually, all those involved in 

preparing this Manual. Most of them are my fellow professionals and well-respected within the 

profession. I wish to record my sincere thanks and appreciation to all of them and I am confident 

that their contributions will be truly appreciated by the readers for many years to come. 

Dato’ Ir. Hj. Ahmad Hussaini bin Sulaiman, 

Director General, 

Department of Irrigation and Drainage Malaysia 

ii March 2009


Table of Contents 

Disclaimer .................................................................................................................................. i 

Foreword .................................................................................................................................. ii 

Table of Contents ...................................................................................................................... iii 

List of Volumes ........................................................................................................................ iv 

Part 1 

GEOTECHNICAL MANUAL 

Part 2 

SITE INVESTIGATION 

Part 3 

ENGINEERING SURVEY 

March 2009 

iii


List of Volumes 

Volume 1 











FLOOD MANAGEMENT 

RIVER MANAGEMENT 

COASTAL MANAGEMENT 

HYDROLOGY AND WATER RESOURCES 

IRRIGATION AND AGRICULTURAL DRAINAGE 

GEOTECHNICAL MANUAL, SITE INVESTIGATION AND ENGINEERING SURVEY 

ENGINEERING MODELLING 

MECHANICAL AND ELECTRICAL SERVICES 

DAM SAFETY 

CONTRACT ADMINISTRATION 

CONSTRUCTION MANAGEMENT 

iv March 2009


Acknowledgements 

Steering Committee: 

Dato’ Ir. Hj. Ahmad Husaini bin Sulaiman, Dato’ Nordin bin Hamdan, Dato’ Ir. K. J. Abraham, Dato’ 

Ong Siew Heng, Dato’ Ir. Lim Chow Hock, Ir. Lee Loke Chong, Tuan Hj. Abu Bakar bin Mohd Yusof, 

Ir. Zainor Rahim bin Ibrahim, En.Leong Tak Meng, En. Ziauddin bin Abdul Latiff, Pn. Hjh. Wardiah 

bte Abd. Muttalib, En. Wahid Anuar bin Ahmad, Tn. Hj. Zulkefli bin Hassan, Ir. Dr. Hj. Mohd. Nor bin 

Hj. Mohd. Desa, En. Low Koon Seng, En.Wan Marhafidz Shah bin Wan Mohd. Omar, Ir. Md Fauzi bin 

Md Rejab, En. Khairuddin bin Mat Yunus, Cik Khairiah bt Ahmad, 

Coordination Committee: 

Dato’. Nordin bin Hamdan, Dato’ Ir. Hj. Ahmad Fuad bin Embi, Dato’ Ong Siew Heng, Ir. Lee Loke 

Chong, Tuan Hj. Abu Bakar bin Mohd Yusof, Ir. Zainor Rahim bin Ibrahim, Ir. Cho Weng Keong, En. 

Leong Tak Meng, Dr. Mohamed Roseli Zainal Abidin, En. Zainal Akamar bin Harun, Pn. Norazia 

Ibrahim, Ir. Mohd. Zaki, En. Sazali Osman, Pn. Rosnelawati Hj. Ismail, En. Ng Kim Hoy, Ir. Lim See 

Tian, Ir. Mohd. Fauzi bin Rejab, Ir. Hj. Daud Mohd Lep, Tn. Hj. Muhamad Khosim Ikhsan, En. Roslan 

Ahmad, En. Tan Teow Soon, Tn. Hj. Ahmad Darus, En. Adnan Othman, Ir. Hapida Ghazali, En. 

Sukemi Hj. Sidek, Pn. Hjh. Fadzilah Abdul Samad, Pn. Hjh. Salmah Mohd. Som, Ir. Sahak Che 

Abdullah, Pn. Sofiah Mat, En. Mohd. Shafawi Alwi, En. Ooi Soon Lee, En. Muhammad Khairudin 

Khalil, Tn. Hj. Azmi Md Jafri, Ir. Nor Hisham Ghazali, En. Gunasegaran M., En. Rajaselvam G., Cik Nur 

Hareza Redzuan, Ir. Chia Chong Wing, Pn Norlida Mohd. Dom, Ir. Lee Bea Leang, Dr. Hj. Md. Nasir 

Md. Noh, Pn Paridah Anum Tahir, Pn. Nurazlina Mohd Zaid, PWM Associates Sdn. Bhd., Institut 

Penyelidikan Hidraulik Kebangsaan Malaysia (NAHRIM), RPM Engineers Sdn. Bhd., J.U.B.M. Sdn. Bhd. 

Working Group: 

Pn. Rozaini binti Abdullah, En. Azren Khalil, Tn. Hj Fauzi Abdullah, En. Che Mohd Dahan Che Jusof, 

En. Ng Kim Hoy, En. Dzulkifli bin Abu Bakar, Pn. Che Shamsiah bt Omar, En. Mohd Latif Bin Zainal, 

En. Mohd Jais Thambi Hussein, En. Osman Mamat, En. Tajudin Sulaiman, Pn. Rosilawani binti 

Sulong, En. Ahmad Solihin Budarto, En. Noor Azlan bin Awaludin, Pn. Mazwina bt Meor Hamid, En. 

Muhamad Fariz bin Ismail, Cik Sazliana bt Abu Omar, Cik Saliza Binti Mohd Said, En. Jaffri Bahan, En. 

Mohd Idrus Amir, Mej (R) Yap Ing Fun, Ir Mohd Adnan Mohd Nor, Ir Liam We Lin, Ir. Steven Chong, 

En. Jamal Abdullah, En. Ahmad Ashrin Abdul Jalil, Cik Wan Yusnira Wan Jusoh @ Wan Yusof. 

March 2009 

i


Registration of Amendments 

Amend 

No 

Page 

No 

Date of 

Amendment 

Amend 

No 

Page 

No 

Date of 

Admendment 

ii March 2009



Acknowledgements ..................................................................................................................... i 

Registration of Amendments ...................................................................................................... ii 


List of Symbols ......................................................................................................................... iv 

Chapter 1 

Chapter 2 

Chapter 3 

Chapter 4 

Chapter 5 

Chapter 6 

Chapter 7 

Chapter 8 

Chapter 9 

Chapter 10 

GENERAL 

GEOTECHNICAL DESIGN PROCESS 

FUNDAMENTAL PRINCIPLES 

SOIL SETTLEMENT 

BEARING CAPACITY THEORY 

SLOPE STABILITY 

RETAINING WALL 

GROUND IMPROVEMENT 

FOUNDATION ENGINEERING 

SEEPAGE 

March 2009 

iii


List of Symbols 

γ 

Unit weight 

γ d Dry unit weight 

γ w Unit weight of water 

γ b 

S 

w 

e 

e 0 

n 

G 

σ 

u 

s 

Buoyant unit weight 

Degree of saturation 

Moisture content 

Void ratio 

Initial void ratio 

Porosity 

Specific gravity of solids 

Total stress 

Pore water pressure 

σ’ Effective stress 

g 

Gravity 

ρ w 

c 

C c 

C r 

U 

t 

θ 

δ 

q ult 

q u 

 

Density of water 

Cohesion 

Compression Index 

Recompression Index 

Degree of consolidation 

Time 

Angular distortion 

Differential settlement in the structure 

Ultimate net bearing capacity 

Allowable net bearing capacity 

Frictional angle 

’ Effective frictional angle 

K a 

K p 

E s 

Coefficient of active earth pressure 

Coefficient of passive earth pressure 

Young’s modulus of soil 

iv March 2009

PART 1: GEOTECHNICAL MANUAL

CHAPTER 1 GENERAL

Chapter 1 GENERAL 


Table of Contents ......................................................................................................... 1-i 

1.1 PURPOSE AND SCOPE ....................................................................................... 1-1 

1.2 LIMITATION OF MANUAL ................................................................................... 1-1 

March 2009 1-i


(This page is intentionally left blank) 

1-ii March 2009


1 GENERAL 

1.1 PURPOSE AND SCOPE 

Part 1 Volume 6 is developed around the aspects of geotechnical engineering usually required in 

JPS nature of work, that include earth retaining structures, river works, embankment, revetment, 

slope stability and stabilization works as well as the various coastal and hydraulic related works. It 

serves to provide a very selective and by no means comprehensive overview of fundamental 

practical knowledge ranging from methods of theoretically based analysis to “rules of thumb” 

solutions for geotechnical and foundation analysis, design and construction issues encountered in 

JPS work. 

It is envisaged that this manual will most likely be used by practicing civil generalists, geotechnical 

and foundation specialists, and others involved in the planning, design and construction of JPS’s 

nature of works. 

The main goals of this Part are to:- 

a) Provide a general understanding and appreciation of the geotechnical principles gearing 

towards a sound, safe and cost-effective design and construction of JPS projects. 

b) Serve as a consistent guidance for the practitioners involved in the geotechnical planning, 

design and construction in all phases of a JPS project. 

c) Encourage the readers to follow through the topic of interest in one or more of the 

reference books mentioned in the references 

1.2 LIMITATION OF MANUAL 

Even though the material presented is theoretically correct and represents the current state-of-thepractice, 

the user must realize that there is no possible way to cover all the various intricate aspects 

of geotechnical engineering. Owing to the high degree of ambiguities and uncertainties in the 

various aspect of geotechnical engineering, sound engineering judgment from highly experience 

and competent specialist practicing engineer is most important. For example, the values for the 

parameters to be used in the analysis and design should be selected by a geotechnical specialist 

who is intimately familiar with the type of soil in that region and intimately knowledgeable about 

the regional construction procedures that are required for the proper installation of such 

foundations in local soils. Often the key in the successful practice and application of geotechnical 

engineering lies in a sound knowledge and understanding of the engineering properties and 

behavior of soils in situ when subjected to changes in the environment conditions such as 

engineering loading or unloading. 

March 2009 1-1



1-2 March 2009

CHAPTER 2 GEOTECHNICAL DESIGN PROCESS

Chapter 2 GEOTECHNICAL DESIGN PROCESS 


Table of Contents .................................................................................................................. 2-i 

List of Tables ....................................................................................................................... 2-ii 

List of Figures ...................................................................................................................... 2-ii 

2.1 GENERAL ................................................................................................................. 2-1 

2.2 DESIGN PROCESS ..................................................................................................... 2-1 

2.2.1 Determine Type of Geotechnical Design and Parameters Required ................. 2-2 

2.2.2 Decide on Appropriate Geotechnical Investigation ......................................... 2-5 

2.2.3 Interpret Geotechnical Investigation Result to Obtain Representative 

Parameters/Properties ................................................................................ 2-5 

2.2.4 Designer’s Analysis and Design ................................................................... 2-6 

2.2.5 Check Compliance and Need for Modification during Construction .................. 2-6 

2.2.6 Post Construction Monitoring and Verification of Structure Performance .......... 2-7 

REFERENCES ....................................................................................................................... 2-8 

March 2009 2-i


List of Tables 

Table Description Page 

2.1 Typical Scope of DID Works (After Geotechnical Guidelines for DID Works) 2-3 

2.2 Type Of Geotechnical Analysis Corresponding To Design Component 2-3 

List of Figures 

Figure Description Page 

2.1 Flow Chart for the Designer Involvement in Geotechnical Design 2-2 

2.2 Some Typical DID's Structures 2-4 

2.3 Combination of Sources of Information in Geotechnical Design 2-6 

2-ii March 2009


2 GEOTECHNICAL DESIGN PROCESS 

2.1 GENERAL 

Geotechnical engineering is highly empirical and is perhaps much more of an ‘art’ than the other 

disciplines within civil engineering because of the basic nature of soil and rock materials. They are 

often highly variable, heterogeneous and anisotropic i.e. their engineering and material properties 

may vary widely within the soil mass and also may not be the same in all direction. Furthermore, 

the behavior of soil and rock materials are often controlled by the joints, fractures, weak layers and 

zones and other ‘defects’ in the materials. 

In the application of geotechnical engineering, the soil is usually assumed to be homogenous and 

isotropic obeying linear stress-strain laws. However, to account for the real material behavior, large 

empirical correction or ‘factors of safety’ must be applied in geotechnical design. As such, 

geotechnical engineering is really an ‘art’ rather than an engineering science, where good judgment 

and practical experience of the designer and contractors are essential for a successful geotechnical 

design. 

2.2 DESIGN PROCESS 

In geotechnical engineering, the analysis and design process normally involved the various steps as 

illustrated in Figure 2.1. It includes determination of the type of geotechnical design and their 

required parameters, identification of appropriate geotechnical investigation works, evaluation and 

interpretation of geotechnical investigation result to obtain representative parameters and 

properties, performing design and analysis, checking compliance during construction and post 

construction monitoring. 

March 2009 2-1


DESIGNER ASSIGNED PROJECT 

DETERMINE TYPE OF GEOTECHNICAL DESIGN 

AND PARAMETERS REQUIRED 

DECIDE ON APPROPRIATE GEOTECHNICAL 

INVESTIGATIONS 

INTERPRET GEOTECHNICAL INVESTIGATION RESULT TO 

OBTAIN REPRESENTATIVE PARAMETERS/PROPERTIES 

DESIGNER’S ANALYSIS AND DESIGN 

CHECK COMPLIANCE AND NEED FOR MODIFICATION 

DURING CONSTRUCTION 

POST CONSTRUCTION MONITORING AND VERIFICATION OF 

STRUCTURE PERFORMANCE 

Figure 2.1 Flow Chart for the Designer Involvement in Geotechnical Design 

2.2.1 Determine Type of Geotechnical Design and Parameters Required 

The type of geotechnical analysis and design depends very much on the type of structures or works 

to be designed. Table 2.1 below highlighted the types of works normally carried out by DID and 

their associated design components which include various hydraulic structures; embankments and 

dams; subsurface drainage; excavations; earth retaining structures and revetment works. The type 

of geotechnical analysis required and corresponding to the design components are as in Table 2.2, 

namely bearing capacity, settlement, slope stability, seepage, retaining wall, soil and geosynthetic 

filter. 

2-2 March 2009


Table 2.1 Typical Scope of DID Works (After Geotechnical Guidelines for DID Works) 

Design 

Components 

Scope of 

Work 

1. River Works 

and Erosion 

control 

2. Irrigation and 

Drainage 

3. Flood 

Mitigation 

Hydraulic 

Structure 

Embankments 

and Dams 

Sub-surface 

Drainage 

Excavation 

Works 

X X X X 

X X X X 

Retaining 

Structures 

Revetment 

X X X 

4. Urban Drainage X X X X X X 

5. Coastal 

Engineering 

X X X 

Table 2.2 Type Of Geotechnical Analysis Corresponding To Design Component 

Geotechnical 

Analyses 

Design 

Components 

1. Hydraulic 

Structure 

2. Embankments 

and Dams 

3. Retaining 

Structure 

4. Subsurface 

Drainage 

Bearing 

Capacity 

Settlement 

Slope 

Stability 

Seepage 

Retaining 

wall 

Soil and 

Geosynthetic 

Filter 

X X X X X 

X X X X 

X X X X X 

X X X 

5. Excavations X 

6. Revetments X X X 

Some typical DID structures are as shown in Figure 2.2 

March 2009 2-3


Figure 2.2 Some Typical DID's Structures 

2-4 March 2009


2.2.2 Decide on Appropriate Geotechnical Investigation 

The objectives and various general details on the type of geotechnical investigation works are 

described in Part 2, Volume 6 : Soil Investigation which include both field and laboratory works. 

Suffice here to mention that the composition and amount of geotechnical investigation proposed 

shall be able to provide sufficient data on the ground, groundwater conditions at the proposed site 

and proper description of the essential soil properties for geotechnical design and construction. It 

shall also be planned to take into account the construction and performance requirements of the 

proposed structure. 

Very often geotechnical engineer is required to determine the type of soil investigation works in 

relation to the envisage analysis required in the design works, i.e. the long-term (drained with 

effective stress analysis) or short-term analysis (undrained total stress analysis) conditions. 

2.2.3 Interpret Geotechnical Investigation Result to Obtain Representative 

Parameters/Properties 

The evaluation and interpretation of geotechnical investigation work shall include a review of the 

field and laboratory results to derive at the reasonable and representative parameters and 

properties. This normally involves tabulation and graphical presentation of field and laboratory 

results such as the range and distribution of values of the required soil parameters (including 

ground water condition), subsurface strata profile which differentiate and group the various 

formations and properties. Any irregularities or adverse field and laboratory results shall be pointed 

out, commented upon, and if necessary to propose further geotechnical investigation for 

verification. Reader should refer to Part 2 Volume 6 for more detail and comprehensive information 

on this topic. 

In spite of the many advances in geotechnical engineering theory, there are still many uncertainties 

in the analysis and design due mainly to the highly variable, heterogeneous and anisotropic nature 

of soil material. Designer normally use various investigation and testing techniques to determine the 

soil conditions, however even the most thorough investigation program encounters only a small 

portion of the soils and relies heavily on the interpolation and extrapolation. The most practical 

approach to solve geotechnical design issues is to combine the sources of information gathered 

through soil investigation and testing program, established theory developed to predict the behavior 

of soils and experience obtained from previous projects coupled with sound engineering judgment. 

These approaches are depicted in Figure 2.3 

March 2009 2-5


Site 

Investigation/ 

laboratory 

Testing 

Established 

Theory 

Experience 

and Judgment 

Figure 2.3 Combination of Sources of Information in Geotechnical Design 

2.2.4 Designer’s Analysis and Design 

Some the common geotechnical analysis and design carried out by the Department include 

evaluation and determination of the soil bearing capacity, settlement, seepage forces; and stability 

of slope, earth retaining structures as well as the selection of effective soil and geosynthetic filter in 

sub-soil drainage. 

In carrying out the analysis and design, sound engineering by experience geotechnical engineer 

should be incorporated to compensate for the many uncertainties in actual soil behavior, which 

should take into consideration the following factors: 

• Required reliability or acceptable probability of failure 

• Consequence of failure 

• Degree of uncertainties in soil properties and applied loads 

• Compromise between cost and reliability 

• Degree of ignorance of the structure behaviour 

2.2.5 Check Compliance and Need for Modification during Construction 

During construction, site operation shall be checked for compliance with the method of construction 

assumed in the design. Also, observation and measurements of the structure and its surrounding 

may necessitate some remedial measures or alterations to the construction sequence, for example 

the unexpected excessive settlement of the embankment under construction would warrant the 

review of the design and proposed sequence of construction. In fact, a great deal of geotechnical 

information can be gathered during construction phase of a project, particularly those involving 

huge volume of earth excavation or exposure where the actual ground conditions can be identified. 

These information should then be used to validate the geotechnical design assumptions or soil 

parameters and if necessary, to revise and modify the design accordingly. 

2-6 March 2009


2.2.6 Post Construction Monitoring and Verification of Structure Performance 

A geotechnical design should not be considered completed upon the completion of the construction 

works. The designer should also be involved in post-construction activities such as visual 

observation and inspection of the structure; gathering and analyzing results of instrumentation 

monitoring to ensure its long-term performance and identified any necessary maintenance work. 

Any lesson learned from the design stage to the completion of the construction works should be 

adequately documented for future references. 

March 2009 2-7


REFERENCES 

[1] Bowles, J.E. Foundation Analysis and Design. (Fourth edition). McGraw-Hill International, 

New York, 1992, 1004 p. 

[2] Brown, R.W., (1996) Practical foundation Engineering Handbooks, Mcgraw-Hill 

[3] BSI. Eurocode 7: Geotechnical Design – Part 1: General Rules (BS EN 1997-1 : 2004). British 

Standards Institution, London, 2004, 117 p. 

[4] Carter M. & Symons, M.V., Site Investigations and foundations Explained, Pentech Press, 

London 

[5] CGS, Canadian Foundation Engineering Manual, (Third edition). Canadian Geotechnical 

Society, Ottawa, 1992, 512 p. 

[6] Das, B.M., Principles of Geotechnical Engineering, PWK-Kent Publishing Company , 

Boston,MA., 1990 

[7] Dept. of the Navy, Bureau of Yards and Docks, Washington D.C., NAVFAC DM-7.1, May 

1982, Soil Mechanics 

[8] DID, Geotechnical Guidelines for D.I.D Works 

[9] Holtz, R.D., Kovacs, W.D. An Introduction to Geotechnical Engineering, Prentice-Hall, Inc. 

New Jersey 

[10] Koerner R.M .• Construction and Geotechnical Method in Foundation Engineering, McGraw 

Hill, 1985. 

[11] Lambe T.W. and Whitman R.V., Soil Mechanics, John Wiley 8: Sons, 1969 

[12] Peck R.B Hanson W.E. and Thornburn R.H., “Foundation Engineering", John Wiley and Sons, 

1974. 

[13] Smith C.N., Soil Mechanics for Civil and Mining Engineers. 

[14] Teng W.C., Foundation Design, Prentice Hall, 1984. 

[15] Terzaghi, K. & Peck, R.B. (1967). Soil Mechanics in Engineering Practice. (Second edition). 

Wiley, New York, 729 p. 

2-8 March 2009

CHAPTER 3 FUNDAMENTAL PRINCIPLES

Chapter 3 FUNDAMENTAL PRINCIPLES 


Table of Contents .................................................................................................................... 3-i 

List of Tables ......................................................................................................................... 3-ii 

List of Figures ........................................................................................................................ 3-ii 

3 FUNDAMENTAL PRINCIPLES ................................................................................................. 3-1 

3.1 BASIC WEIGHT-VOLUME RELATIONSHIPS ..................................................................... 3-1 

3.2 EFFECTIVE STRESS CONCEPT ....................................................................................... 3-2 

3.3 VERTICAL STRESS DISTRIBUTION ................................................................................ 3-4 

3.4 SHEAR STRENGTH ....................................................................................................... 3-5 

3.4.1 Basic Principle................................................................................................. 3-5 

3.4.2 Effective Versus Total Stress Analysis ............................................................... 3-8 

REFERENCES ....................................................................................................................... 3-11 

March 2009 3-i




3.1 Definition and Typical Values of Common Soil Weight-Volume Parameters 3-1 

3.2 Some Unit Weight Volume Inter-Relationships 3-2 

3.3 Design Conditions and Related Shear Strengths and Pore Pressures 3-10 



3.1 Unit Soil Mass and Phase Diagram 3-1 

3.2 Total Stress at a Point 3-2 

3.3 Example 3.1 3-3 

3.4 Schematic of the Vertical Stress Distribution with Depth under an Embankment generated 

by FoSSA Program (from Soil and Foundation - FHWA) 3-4 

3.6 Graphical Representative of Shear Strength 3-7 

3.7 Mohr-Coulomb’s Circles and Failure Envelopes 3-8 

3-ii March 2009


3 FUNDAMENTAL PRINCIPLES 

3.1 BASIC WEIGHT-VOLUME RELATIONSHIPS 

Soil mass is generally idealized as a three phase system consisting of solid particles, water and air 

as illustrated in diagram in Figure 3.1. Owing to the three different components of soils, complex 

states of stresses and strains may exist in a soil mass. The various volume changes phenomena 

encountered in geotechnical engineering, such as deformation, consolidation, collapse, compaction, 

expansion, shrinkage etc. can be described in term of the various volumes of these components in 

the soil mass. Thus, knowledge of the relative proportion of each component and their various 

inter-relationships can give an important insight into engineering behavior of a particular soil. 

The weight-volume relationships of the soil mass are readily available in most soil mechanics 

textbooks. Most of these relationships are as summarized in Table 3.1 and Table 3.2. 

Soil particles 

Volume 

V a 

Air 

Weight 

W a ≈0 

Voids (filled with 

water and air) 

V 

V v 

V w 

V s 

Water 

Solid 

W w 

W s 

W 

1 unit 

Figure 3.1 Unit Soil Mass and Phase Diagram 

Table 3.1 Definition and Typical Values of Common Soil Weight-Volume Parameters 

Typical Range 

Parameter Symbol Definition English SI 

W 

Unit weight 

 

90 – 130 lb/ft 3 14 – 20 kN/m 3 

V 

W 

Dry unit weight 

s 

d 

60 – 125 lb/ft 3 9 – 19 kN/m 3 

V 

W 

Unit weight of water 

w 

w 

62.4 lb/ft 3 9.8 kN/m 3 

V 

Buoyant unit weight b sat - w 28 – 68 lb/ft 3 4 – 10 kN/m 3 

Degree of saturation 


Void ratio 

Porosity 

Specific gravity of solids 

(Source: Donald P. Coduto, [6]) 

S 

w 

e 

n 

G s 

V w 

V v 

x 100% 2 – 100% 2 – 100% 

W w 

x 100% 

W s 

3 – 70% 3 – 70% 

V v 

V s 

0.1 – 1.5 0.1 – 1.5 

V v 

V x 100% 9 – 60% 9 – 60% 

W s 

V s w 

2.6 – 2.8 2.6 – 2.8 

March 2009 3-1


Table 3.2 Some Unit Weight Volume Inter-Relationships 

Unit-weight Relationship Dry Unit Weight (No Water) Saturated Unit Weight (No Air) 

t = 1+wG s w 

1+e 

t 

d = 

1+w 

 

sat 

= G s+e w 

1+ 

e 

t = G s+Se w 

1+e 

t = 1+wG s w 

1+ wG s 

S 

t =G s w 1-n(1+w) 

d = G s t 

1+e 

d =G s w (1-n) 

t = 

G s w 

1+ wG s 

S 

= 

eS w 

d 

1+ 

ew 

dsa 

t ‐n w 

d = sat - e 

1+e w 

sat 1-nG s +n w 

sat = 1+w 

1+wG s 

G s w 

sat e w 1+w 1+e w 

 

n 

sat d w 

sat d e 

1+e w 

In above relations, w refers to the unit weight of water, 62.4 pcf (=9.81 kN/m 3 ). 

(Source: Donald P. Coduto, [6]) 

3.2 EFFECTIVE STRESS CONCEPT 

The concept of effective stress was first proposed by Karl Terzaghi in the mid sixties. It is a simple 

concept with significant implications on how the science of geotechnical engineering develops. In 

simple terms the concept stipulates that soil consists of 2 major components in general, i.e., (i) 

particulate, and (ii) pore water. 

Under an applied load, the total stress (σ) in a saturation unit soil mass is composed of intergranular 

stress and the pore water pressure (u) as illustrated in Fig 3.2. When pore water drains 

from the soil, the contact between the soil grains will increase which increases the inter-granular 

stress. The inter-granular stress is called the effective stress, σ’. 

Particles 

Pore Water 

Mathematically, 

σ = σ’ + u 

Where σ = Total stress 

σ' = effective stress 

u = pore water pressure 

Figure 3.2 Total Stress at a Point 

The concept of effective stress is extremely useful in the development of soil strength theories and 

soil behaviour models. It allows a better understanding of soil behaviour, interpreting laboratory 

test results and making engineering design calculations such as in the estimation of settlement due 

to consolidation. More significantly, the concept implies that the soil shearing strength depends only 

on the effective stress componentpore water carries no shear under hydrostatic or steady state 

seepage conditions (i.e., flow velocity is negligible). 

3-2 March 2009


Both the total stress and pore water pressure may readily be estimated or calculated with 

knowledge of the densities and thickness of soil layers and location of ground water stable. To 

calculate the total vertical stress σ v at a point in a soil mass, you simply sum up the weights of all 

the material (soil solids + water) above that point multiplied by respective thickness of each soil 

layer or 

n 

ρ i 

σ v = ∑i= gz i (3.1) 

σ v = Vertical stress 

ρ i = Densities of each layer above point in question 

g = Gravity 

z = Thickness of each layer 

n = Number of layers above point in question 

The pore water pressure is similarly calculated for static water conditions i.e. 

u = ρ w g z w (3.2) 

Where ρ w = density of water 

z w = depth below ground water table to the point in question 

Example: 3.1 

Given that the container of soil shown in Fig 3.3 with the saturated density as 2.0 Mg/m 3 

Calculate the total and effective stress at Elevation A 

Water 

Z w = 2 m 

Soil 

h = 5 m 

Elev. A 

Figure 3.3 Example 3.1 

The stresses at Elevation A due to the submerged soil and water above are: 

Total stress = ρ sat g h + ρ w g z w 

= (2 x 9.81 x 5.0) + (1 x 9.81 x 2.0) 

= 117.7 kPa 

Pore water pressure, u 

= ρ w g (z w + h) 

= 1 x 9.81 x (2 + 5) 

= 68.7 kPa 

Effective stress at Elev. A, σ ’ 

= σ − u = ( ρ sat g h + ρ w g z w ) - ρ w g (z w + h) 

= 117.7 - 68.7 = 49.0 kPa 

March 2009 3-3


3.3 VERTICAL STRESS DISTRIBUTION 

When a very large area is to be loaded, the induced stress in underneath soil would be would be 

100% of the applied stress at the contact surface. However, near the edge or end of the loaded 

area you might expect a certain amount of attenuation of stress with depth because no stress is 

applied beyond the edge. Likewise, with a footing of limited size the applied stress would dissipate 

rather rapidly with depth. 

Figure 3.4 illustrated a schematic of the vertical stress distribution with depth along the center line 

under an embankment of height, h, constructed with a soil having total unit weight, γ t . 

Figure 3.4 Schematic of the Vertical Stress Distribution with Depth under an Embankment 

generated by FoSSA Program (from Soil and Foundation - FHWA) 

One of the simplest methods to compute the distribution of stress with depth for a loaded area is to 

use the 2 to 1 (2:1) method. This is an empirical approach based on the assumption that the area 

over which the load acts increases in a systematic way with depth. Since the same vertical force is 

spread over an increasingly larger area, the unit stress decreases with depth, as shown in Fig. 3.4. 

In Fig. 3.5a, a strip or continuous footing is seen in elevation view. At a depth z, the enlarged area 

of the footing increases by z/2 on each side. The width at depth z is then B + Z and the stress σ z 

at that depth is 

σ z = 

load 

B+z×1 = σ o(B×1) 

(B+z)×1 

(3.3) 

By analogy, the corresponding stress at depth z for a rectangular footing of width B and length L 

(as illustrated in Figure 3.5b would be 

∆σ z = 

load 

B+z(L+z) = 

σ o BL 

B+z(L+z) 

(3.4) 

3-4 March 2009

Chapter 3 FUNDAMENTAL 

PRINCIPLES 

Figure 3.5 The 2:1 Method for Estimation of Vertical Stress Distribution with Depth 

3.4 

3.4.1 

SHEAR STRENGTH 

Basic 

Principle 

The shear strength 

of soils is a most important aspect of geotechnical engineering. The bearing 

capacity of shallow or deep foundations, slope stability, retaining wall design are all affected 

by the 

shear strength of the soil. The shear strength of a soil can be defined as the ultimate or maximum 

shear stress the soil can withstand. Geotechnical failure occurs when shear stress inducedd by the 

applied 

loads exceed the shear strength of the soil. 

March 2009 

3-5


The shear strength of soil can be may be expressed by Coulomb’s equation: 

where 

s = c + σ tan φ (3.5) 

s = shear strength or shear resistance 

c = cohesion 

φ = angle of internal friction of soil 

σ = total normal stress to shear plane 

For effective stresses the shear strength is expresses as: 

where 

s = c '+ σ' tan φ' and (3.6) 

σ' = (σ − u) (3.7) 

c' = effective cohesion 

φ' = effective angle of internal friction 

σ' = effective stress or inter-granular stress normal to the shear plane 

u = pore water pressure on the shear plane 

The equation 3.1 and 3.2 could also be represented graphically in Figure 3.6. 

As expressed in the above equations, the shear strength of soil is represented by the additive of 

two terms i.e. σ tan φ (οr σ'tan φ) and c (or c’). The first term is the inter-granular frictional 

component which is approximately proportional to the normal stress on the surface, σ (or σ'), 

whereas the second term is due to the internal electro-chemical bonding between particles and is 

independent of the normal stress. 

A coarse-grained soil such as sand and gravels has no cohesion and thus, it strength depends solely 

on the inter-granular friction between soil grains. This type of soil is called granular, cohesionless, 

non-cohesive or frictional soil. On the other hand, soils containing large amounts of fine grains 

(clay, silt and colloid) are called fine-grained or cohesive soils. 

3-6 March 2009


Figure 3.6 Graphical Representative of Shear Strength 

The shear strength parameters, c and σ or c' and σ ', are normally determined from laboratory 

shear test results such as triaxial and direct shear tests. A series of tests are usually carried out 

whereby the stresses (normal and shear stresses) from each test representing failure are plotted. 

The resulting graph, as illustrated in Figure 3.7, is known as the Mohr-Coulomb (M-C) failure 

envelope which represents the shear strength of the soil. 

March 2009 3-7


M-C Failure Envelope 

M-C Failure Envelope 

Figure 3.7 Mohr-Coulomb’s Circles and Failure Envelopes 

The physical meaning of the M-C failure envelope may be explained as follows: 

• Every point on the M-C failure envelope represents a combination of normal and shear stress 

that results in failure of the soil, i.e. the limiting state of stress for equilibrium. 

• If the state of stress is represented by a point below the M-C failure envelope then the soil 

will be stable for that state of stress. 

• States of stress beyond the M-C failure envelope cannot exist since failure would have 

occurred before that point could be reached. 

3.4.2 Effective Versus Total Stress Analysis 

It is important to note that the properties of soil and its shear strength in the vicinity of construction 

facility could change with time. As explained in Item 3.2, when the stress in the soil is suddenly 

changed (e.g. due to applied load), the additional stress is initially carried by the pore water 

pressure resulting to what is known as excess pore water pressure. If a foundation consolidates 

slowly, relative to the rate of construction, a substantial portion of the applied load will be carried 

by the pore water, which has no shear strength, and the available shearing resistance is limited to 

the in-situ shear strength. In this case, analysis are carried out using the total stress (undrained) 

analysis. 

3-8 March 2009


In time , the excess pore water pressure will dissipate as result of seepage under consolidation and 

the stress is eventually carried by soil skeleton of the soil and under such condition, analysis using 

the effective (drained) stress analysis is applied. Since shear strength will vary with time, it is 

important for the designer to understand and determine at which point in time i.e. before, during or 

after construction that is critical to the design of the structure. 

As granular or sandy soils are more permeable than cohesive or clayey soils, drainage of excess 

pore pressure in sandy soil occurs much more rapidly. Hence, effective (drained) stress analysis is 

usually necessary for sandy soils. For clayey soil, either a total (undrained) stress analysis or 

effective (drain) stress analysis is required depending on the time considered in relation to the 

duration of construction. 

Effective stress analysis requires the estimation of the drained strength parameters c’, φ’ and pore 

pressures. However, with pure free draining sands, φ = φ’ and c = 0. For total stress analysis, 

undrained parameters typically used are φ = 0 and c determined from in-situ vane shear (for soft 

clay) or undrained unconfined (UU) and consolidated undrained (CIU) triaxials tests. 

In general, depending on the soil compressibility, thickness, permeability, nature of the stress 

applied, and duration of construction, designer usually considers the two conditions listed to 

determine which is more critical in the analysis 

a) At the end of construction, e.g. construction of river embankment in soft clay. Geotechnical 

analysis maybe carried using total stress analysis with undrained shear strength parameters 

or effective stress analysis with drained shear strength parameters 

b) Long-term e.g. construction of pervious reinforced earth retaining structure using free 

draining backfill. Long-term geotechnical analysis is normally carried out using effective 

stress analysis with drained shear strength parameters and estimated or measured pore 

pressures. 

Table 3.3 provided a more detail design conditions in relation to appropriate shear strengths for use 

in analyses of static loading conditions. 

March 2009 3-9


Table 3.3 Design Conditions and Related Shear Strengths and Pore Pressures 

Shear Strengths and Pore Pressures for Static Design Conditions 

Design Condition Shear Strength Pore Water Pressure 

During Construction Free draining soils – use drained Free draining soils – Pore water 

and End-of- 

shear strengths related to pressures can be estimated using 

Construction 

effective stresses 

analytical techniques such as 

hydrostatic pressure computations if 

there is no flow or using steady 

seepage analysis techniques (flow 

nets or finite element analyses). 

Low permeability soils – use 

undrained shear strengths 

related to total stresses 

Low-permeability soils = Total 

stresses are used, pore water 

pressures are set to zero in the slope 

stability computations. 

Steady-State 

Seepage Conditions 

Use drained shear strength 

related to effective stresses. 

Pore water pressures from field 

measurements, hydrostatic pressure 

computations for no-flow conditions, 

or steady seepage analysis techniques 

(flow nets or finite difference 

analyses). 

Sudden Drawdown 

Conditions 

Free draining soils – use drained 

shear strengths related to 

effective stresses. 

Free draining soils – First-stage 

computations (before drawdown) – 

steady seepage pore pressures as for 

steady seepage condition. Secondand 

third-stage computations (after 

drawdown) – pore water pressures 

estimated using same techniques as 

for steady seepage, except with 

lowered water level. 

Low permeability soils – Threestage 

computations: First stage 

– use drained shear strength 

related to effective stresses, 

second stage – use undrained 

shear strengths related to 

consolidation pressures from the 

first stage, third stage – use 

drained strengths related to 

effective stresses, or undrained 

strengths related to 

consolidation pressures from the 

first stage, depending on which 

strength is lower – this will vary 

along the assumed shear 

surface. 

Low-permeability soils – First-stage 

computations – steady state seepage 

pore pressures as described for steady 

seepage condition. Second–stage 

computations – total stresses are 

used, pore water pressures are set to 

zero. Third-stage computations – 

same pore pressures as free draining 

soils if drained strengths are used, 

pore water pressures are set to zero 

where undrained strengths are used. 

3-10 March 2009


REFERENCES 

[1] Bishop A.V and Henkel D.J., The Measurement of Soil Properties in the Triaxial Test, 

E.Arnold, 1962. 


New York, 1992, 1004 p. 





London 

[6] Donald P.Coduto, Foundation Design, Principles and Practices 




1982, "Soil Mechanics" 

[9] Dept. of the Navy, Bureau of Yards and Docks, Washington D.C.,NAVFAC DM-7.2, May 1982, 

"Foundations and Earth Structures" 


New Jersey 

[11] Koerner R.M . Construction and Geotechnical Method in Foundation Engineering, McGraw 

Hill, 1985. 

[12] Ladd C.C., Foott R., Ishihara K., Schlosser F., and Roulos H.G., Stress Deformation and 

Strength Characteristics, State of the Art Report, Session I, IX ICSMFE, Tokyo, Vol. 2, 1971, pp. 421 

- 494. 


[14] McCarthy D.J., Essentials of Soil Mechanics and Foundations. 

[15] Nayak N. V. I II Foundation Design Manual. Dhanpat Rai a Sons I 1982. 

[16] Peck R.B Hanson W.E. and Thornburn R.H., Foundation Engineering, John Wiley and Sons, 

1974. 





March 2009 3-11


[20] U.S. Department of Transportation, Soil and Foundation, Reference Manual Volume 1 & 2 

(2006) 

3-12 March 2009

CHAPTER 4 SOIL SETTLEMENT

Chapter 4 SOIL SETTLEMENT 





4 SOIL SETTLEMENT .............................................................................................................. 4-1 

4.1 GENERAL CONCEPT .................................................................................................... 4-1 

4.1.1 Immediate (Distortion) Settlement ................................................................ 4-1 

4.1.2 Primary Consolidation ................................................................................... 4-2 

4.1.3 Secondary Compression ................................................................................ 4-2 

4.2 SETTLEMENT ON GRANULAR SOILS .............................................................. 4-2 

4.3 ESTIMATION OF PRIMARY CONSOLIDATION IN COHESIVE SOIL .................................... 4-3 

4.3.1 Normally Consolidated Soils .......................................................................... 4-5 

4.3.2 Overconsolidated (Preconsolidated) Soils ....................................................... 4-5 

4.3.3 Underconsolidated Soils ................................................................................ 4-6 

4.4 RATE OF CONSOLIDATION .......................................................................................... 4-7 

4.5 SECONDARY SETTLEMENT OF COHESIVE SOIL ............................................................. 4-9 

4.6 DIFFERENTIAL SETTLEMENT ..................................................................................... 4-10 

4.7 PLATE LOADING TEST FOR SETTLEMENT ESTIMATION ............................................... 4-12 

4.8 SETTLEMENT OF RAFT/MAT FOUNDATIONS ............................................................... 4-12 

REFERENCES ....................................................................................................................... 4-14 

March 2009 4-i




4.1 Typical Allowable Total Settlements for Foundation Design 4-3 

4.2 Typical Values of Tolerable Differential Settlement 4-11 



4.1 Components Of Total Settlement Versus Log Time 4-1 

4.2 Typical e – lop p Curve 4-4 

4.3 Typical Consolidation Curve for Normally Consolidated Soil 4-5 

4.4 Typical Consolidation Curve for Over Consolidated Soil 4-6 

4.5 Typical Consolidation Curve for Under-Consolidated Soil 4-7 

4.6 Average Degree of Consolidation U versus Time Factor, Tv under Various Drainage 

Conditions 4-8 

4.7 Example 4.1 4-9 

4.8 The Building was built partly on filled and partly on original ground, which resulted in 

cracks due to excessive differential settlement 4-10 

4-ii March 2009


4 SOIL SETTLEMENT 

4.1 GENERAL CONCEPT 

In geotechnical engineering, in particular foundation works for structures, engineers are interested 

in how much and how fast soil settlement will occur. Excessive settlement including (differential 

settlement) may cause structural damage as well as impair the functionality or serviceability of the 

structures. 

Soils whether cohesionless or cohesive, will experience settlements immediately after application of 

loads. Whether or not the settlements will continue with time after the application of the loads will 

be a function of how quickly the water can drain from the voids as explained in Item 3.2 Long-term 

consolidation-type settlements are generally not experienced in cohesionless soils where pore water 

can drain quickly or in dry or slightly moist cohesive soils where significant amounts of pore water 

are not present. Therefore, embankment settlements caused by consolidation of cohesionless or 

dry cohesive soil deposits are frequently ignored as they are much smaller compared to immediate 

settlements in such soils. 

The total soil settlement. S t can be divided into 3 main components, namely immediate settlement, 

primary consolidation settlement, , and secondary compression settlement 

S t = S i + S c + S s (4.1) 

S i = immediate settlement 

S c = primary consolidation settlement (time-dependent) 

S s = secondary compression settlement 

S i 

S c 

S s 

Figure 4.1 Components Of Total Settlement Versus Log Time 

4.1.1 Immediate (Distortion) Settlement 

Immediate, or distortion, settlement (S i ) occurs during application of load as excess pore pressure 

develops in the underlying soil. If the soil has a low permeability and it is relatively thick, the excess 

pore pressures are initially undrained. The foundation soil deforms due to the applied shear stresses 

with essentially no volume change, such that vertical compression is accompanied by lateral 

expansion. 

It should be recognized that most field evidence indicates that S i is usually not important design 

consideration especially in cohesive soils. It can usually be reduced by precompression or, to some 

extent, by a controlled loading program which allows consolidation to increase the soil stiffness and 

reduce the shear stress level in the foundation. 

March 2009 4-1


Immediate settlement although not actually elastic is usually estimated by using elastic theory, and 

the procedures for dealing with this problem can be found in textbooks on foundation engineering 

such as Soil and Foundation, FHWA and DID Geotechnical Guidelines. 

4.1.2 Primary Consolidation 

Primary consolidation (S c ) develops with time as drainage allows excess pore pressure to dissipate. 

Volume changes, and thus settlement occur as stresses are transferred from the water (pore 

pressure) to the soil skeleton (effective stress). The rate of primary consolidation is governed by the 

rate of dissipation of pore water pressure. The estimation and rate of primary settlement in 

cohesive soil with low coefficient of permeability are dealt with in more details later in this Chapter. 

4.1.3 Secondary Compression 

Secondary compression settlement (S s ) is the continuing, long term settlement which occurs after 

the excess pore pressures are essentially dissipated and after the effective stresses are practically 

constant. These further volume changes and increased settlements are due to drained creep, and 

are often characterized by a linear relationship between settlement and logarithm of time (refer 

Figure 4.1). 

Secondary compression is normally not very significant relative to the primary consolidation for 

inorganic clayey soil. However, for peats and highly inorganic soils, secondary compression 

constitutes a major part of the total settlement. Reader can refer to Holtz and Kovacs or Soil and 

Foundation, FHWA for guidance on the evaluation of secondary compression settlement. 

4.2 SETTLEMENT ON GRANULAR SOILS 

Most methods for computing the primary settlements of foundations on granular soils are based on 

elastic theory or empirical correlations. Empirical correlations based on standard penetration test 

(SPT) generally provide an acceptable solution for predicting the settlement of a shallow foundation 

on granular soils. 

Poulos (2000) found that although soil behaviour is generally non-linear and highly dependent on 

effective stress level and stress history and hence should be accounted for in settlement analysis, 

the selection of geotechnical parameters, such as the shear and Young's modulus of soils, and site 

characterisation are more important than the choice of the method of analysis. Simple elasticitybased 

methods are capable of providing reasonable estimates of settlements. 

Based on elastic theory, the settlement, δf, of a shallow foundation can be calculated using an 

equation of the following general form: 

δ f = q net B f'f 

E s 

(4.2) 

where 

q net 

B f ' 

Es 

f 

= mean net ground bearing pressure 

= effective width of the foundation 

= Young’s modulus of soil 

= a coefficient whose value depends on the shape and dimensions of the foundation, 

the variation of soil stiffness with depth, the thickness of compressible strata, 

Poisson’s ratio, the distribution of ground bearing pressure and the point at which 

the settlement is calculated. 

4-2 March 2009


Poulos & Davis (1974) gave a suite of elastic solutions for determining the coefficient 'f' for various 

load applications and stress distributions in soils and rocks. 

The increase of stress in soils due to foundation load can be calculated by assuming an angle of 

stress dispersion from the base of a shallow foundation. This angle may be approximated as a ratio 

of 2 (vertical) to 1 (horizontal) (Bowles, 1992; French, 1999). The settlement of the foundation can 

then be computed by calculating the vertical compressive strains caused by the stress increases in 

individual layers and summing the compression of the layers. 

A time correction factor has been proposed by Burland & Burbidge (1985) for the estimation of 

secondary settlement. Terzaghi et al (1991) also give an equation for estimating secondary 

settlement in a similar form. The commencement of secondary settlement is assumed to commence 

when the primary settlement completes, which is taken as the end of construction. 

4.3 ESTIMATION OF PRIMARY CONSOLIDATION IN COHESIVE SOIL 

From the types of settlement described above, generally the most significant settlement is 

consolidation settlement. Consolidation settlement is time dependence. For low permeability soil 

with reasonably thickness, the primary consolidation may take very long time e.g., exceeding 10 

years. Therefore, improvement method by shortening the consolidation process is essential to avoid 

distresses or failure due differential settlement after construction. 

Table 4.1 Typical Allowable Total Settlements for Foundation Design 

Type of Structure 

Typical Allowable Total Settlement, δ a 

(in) 

(mm) 

Office Buildings 

0.5 – 2.1 (1.0 is the most 12 – 50 (25 is the most 

common value) 

common value) 

Heavy Industrial Buildings 1.0 – 3.0 25 – 75 

Bridges 2.0 50 

(Source: Donald P.Coduto [19]) 

In general, lowering of the ground water table will leads to settlement of the ground. In finegrained 

soils, prolonged lowering of water table will cause an increase in the effective stresses by 

extrusion of water from the voids leading to ground settlement. 

Primarily Consolidation, S c (herein refer as ‘consolidation’) is a process when sudden application of a 

load to a saturated soil produces an immediate increase in pore water pressure. Over time, the 

excess pore water pressure will dissipate, the effective stress in the soil will increase and settlement 

will increase. Since shear strength is related to effective stress, it may be necessary to control the 

rate of construction to avoid a shear failure. The rate at which the excess water pressure dissipates, 

and settlement occurs, depends on the permeability of the soil, the amount of water to be expelled 

and the distance the water must travel (drainage path). 

The determination of consolidation is commonly based on the one-dimensional laboratory 

consolidation test results. Typically, the results are expressed in an e-log p plot which is the socalled 

“consolidation curve”, an example of which is as shown as in Figure 4.2. The followings 

parameters r may be obtained from the consolidation curve: 

a) Initial void ratio, eo 

b) Compression index, Cc 

c) Recompression index, Cr 

d) Preconsolidation pressure, p c 

March 2009 4-3

Chapter 


p c 

Figure 4.2 Typical e – lop p Curve 

It should be noted that before this laboratory test results are used, it is very important to 

correct 

the consolidation curves for the 

effects of sampling. The proceduree for correction could be 

readily 

found in most foundation engineering textbooks e.g. Holtz and Kovacs and is not discussed here. 

The response of the soil to settlement also depends on the magnitude of the existing effective 

stress relative to the maximumm past effective stress at a given depth. The overconsolidation ratio, 

OCR, which is a measure of the 

degree of overconsolidation in a soil is defined as 

OCR = pc / 

po 

(4.3) 

where 

pc = preconsolidation pressure (obtained from an e-log p plot) 

po = initial effective vertical stress at the center 

of the layer 

considered. 

The value of OCR provides a basis for determining the effective stress history of 

the clay at the time 

of the proposed loading as follows: 

OCR = 1 : – the clay is considered to be “normally consolidated” under the existing load, i.e., the 

clay has fully consolidated under the existing load (p c = p o ). 

a) 

b) 

OCR > 1 : – the clay is consideredd to be “overconsolidated” under the existing load, i.e., 

the clay has consolidated under a load greater than the load 

that currently exists (pc > p o ). 

OCR < 1 : – the clay is 

consideredd to be “underconsolidated” under the existing load, i.e., 

consolidation under the 

existing load is still occurring and will continue to occur under that 

load until primary consolidation is complete, even if no additional load is 

applied (p c 

4-4 

March 2009


4.3.1 Normally Consolidated Soils 

The settlement of a geotechnical feature or a structure resting on n layers of normally consolidated 

soils (p c = p o) can be computed from Figure 4.3 where n is the number of layers into which the 

consolidating layer is divided: 

c c 

n 

S c = ∑i H o log p f 

10 (4.4) 

1+e 0 p o 

Figure 4.3 Typical Consolidation Curve for Normally Consolidated Soil 

The final effective vertical stress is computed by adding the stress change due to the applied load 

to the initial vertical effective stress. The total settlement will be the sum of the compressions of 

the n layers of soil. 

4.3.2 Overconsolidated (Preconsolidated) Soils 

For overconsolidated clay, i.e., OCR >1, the soils could have in the past subjected to a greater 

stress than exists now. It maybe due to many factors including erosion of the weight of the natural 

soil deosit, removal of the weight of a previously placed fill or structures, etc. 

As a result of preconsolidation, the field state of stress will reside on the initially flat portion of the 

e-log p curve. Figure 4.4 illustrates the case where a load increment, ∆p, is added so that the final 

stress, p f . For this condition, the settlements for the case of n layers of overconsolidated soils will 

be computed by summing the settlements computed from each subdivided compressible layer 

within the zone of influence. 

c c 

n 

S = ∑i (c r log p c 

10 + c c log p f 

10 ) 

1+e 0 p o 

p c 

March 2009 4-5


Figure 4.4 Typical Consolidation Curve for Over Consolidated Soil 

4.3.3 Underconsolidated Soils 

When the state of effective stress of soils has not fully consolidated under an existing load, the soils 

is term as underconsolidation, i.e., OCR < 1. Consolidation settlement due to the existing load, will 

continue to occur under that load until primary consolidation is completed (i.e. under ∆p o ) even if 

no additional load is applied. This condition is represented in Figure 4.5. Thus, any additional load 

increment, ∆p, would have to be added to p o . Consequently, if the soil is not recognized as being 

underconsolidated, the actual total primary settlement due to ∆p o +∆p will be greater than the 

primary settlement computed for an additional load ∆p only, i.e., the settlement may be underpredicted. 

As a result of under-consolidation, the field state of stress will reside entirely on the virgin portion of 

the consolidation curve as shown in Figure 4.5.. The settlements for the case of n layers of underconsolidated 

soils are computed by Equation 4.5 that correspond to Figure 4.5. 

H o 

n 

S = ∑1 (c r log P c 

10 + c c log P f 

10 (4.6) 

1+e o P o P c 

4-6 March 2009


Figure 4.5 Typical Consolidation Curve for Under-Consolidated Soil 

4.4 RATE OF CONSOLIDATION 

The average degree of consolidation, U at any time, t, can be defined as: 

U = S t / S ult (4.7) 

Where S t = Settlement at time of interest 

S ult = Settlement at end of primary consolidation (i.e. at ultimate) when excess pore water 

pressures are zero throughout the consolidating layer 

Figure 4.6 shows the average degree of consolidation (U) corresponding to a normalized time 

expressed in terms of a time factor, T v , where : 

T v = c vt 

H d 

2 

(4.8) 

which can be written 

t T v H d 

2 

C v 

(4.9) 

2 

c v 

= coefficient of consolidation (m /day) 

H d 

= The longest distance to a drainage boundary (m) 

t = time (day) 

March 2009 4-7


Percent consolidation U 

0 

20 

40 

60 

80 

U T v 

10 0.0077 

20 0.0314 

30 0.0707 

40 0.126 

50 0.196 

60 0.286 

70 0.403 

80 0.567 

90 0.848 

100 Infinity 

100 

0 0.2 0.4 0.6 0.8 

Time factor T v 

Figure 4.6 Average Degree of Consolidation U versus Time Factor, Tv under Various Drainage 

Conditions 

Note that the longest drainage distance, H d 

of a soil layer confined by more permeable layers on 

both ends is equal to one-half of the layer thickness. When confined by a more permeable layer on 

one side and an impermeable boundary on the other side, the longest drainage distance is equal to 

the layer thickness. The value of the dimensionless time factor Tv may be determined from Table 

4.6 for any average degree of consolidation. U. The actual time, t, it takes for this percent of 

consolidation to occur is a function of the boundary drainage conditions, i.e., the longest distance to 

a drainage boundary, as indicated by Equation 4.8. By using the normalized time factor, Tv, 

settlement time can be computed for various percentages of settlement due to primary 

consolidation, to develop a predicted settlement-time curve. A typical settlement-time curve for a 

clay deposit under an embankment loading is shown in Figure 4.6 

Coefficient of consolidation, c v can be obtained from laboratory consolidation test data. Two 

graphical procedures are commonly used for this i.e. the logarithm-of-time method (log t) proposed 

by Casagrande and Fadum (1940) and the square-root-of-time method proposed by Taylor (1948). 

These methods are can be found in various textbooks such as Holtz and Kovacs, and Soil and 

Foundations, FHWA. 

4-8 March 2009

Chapter 


Example 4.1: Determine the magnitude of and the time for 90% 

consolidation for the 

settlement of a “wide” embankment as shown in Figure 4.7 

primary 

Figure 4.7 Example 4.1 

a) 

Since the embankment 

is “wide,” the vertical stress at the base of the embankment is 

assumed to 

be the same within the 3 m thick clay layer. Since soil is normally consolidated, 

use Equation 4.3 to determine the primary consolidation settlement as follows: 

b) 

Find the time for 90% consolidationn use Tv = 0. .848 from Figure 4.6. Assume single 

vertical 

drainage due to impervious rock underlying clay 

layer and use Equation 

4.7 to calculate the 

time required for 90% consolidationn to occur. 

4.5 

SECONDARY SETTLEMENT 

OF COHESIVE SOIL 

The traditional method proposed by Buisman (1931) is practical in estimating secondary 

consolidation settlement (Terzaghi et al, 1991; Poulos et al, 2002). In this method, the magnitude 

of secondary consolidation is assumed to vary linearly with the logarithm of time. It is 

usually 

expressed as: 

s c = 

(4.10) 

where 

sc = 

C = 

eo = 

H o = 

t p = 

t s = 

secondary consolidationn 

secondary compression index 

initial void ratio 

Thickness of soils subjecte to secondary consolidation 

time when primary consolidation completed 

time for which secondary consolidation is allowed 

March 2009 

4-9


Mesri et al (1994) proposed correlating the secondary compression index, C , with the compression 

index, C c , at the same vertical effective stress of a soil. He found that the C /C c ration is the 

constant for a soil deposit (see Table 4.2). 

The time at which secondary consolidation is assumed to commence is not well defined. A 

pragmatic approach is to assume that the secondary consolidation settlement commences when 

95% of the primary consolidation is reached (Terzaghi et al, 1991). 

Table 4.2 Values of C /Cc for Geotechnical Materials 

Material 

Granular soil 

Shale and mudstone 

Inorganic clays and silts 

Organic clays and silts 

Peat and muskeg 

(Source: Mesri et al [24]) 

C /Cc 

0.02 ± 0.01 

0.03 ± 0.01 

0.04 ± 0.01 

0.05 ± 0.01 

0.01 ± 0.01 

4.6 DIFFERENTIAL SETTLEMENT 

Damage in structures due to settlement may be classified under 3 categories: 

a) Architectural damage such as cracking in wall partitions and plaster 

b) Structural damage where the structural integrity are affected and 

c) Functional damage where the function of the structure may be impaired. 

Figure 4.8 The Building was built partly on filled and partly on original ground, which resulted in 

cracks due to excessive differential settlement 

Normally, uniform settlement will not give rise to damage. It is the differential settlement that has 

to be controlled. However, differential settlement is difficult to estimate due especially to the nonhomogeneity 

in the ground, and the large variations in the loadings between different supporting 

members. Figure 4.8 illustrates the appearance of crack due to differential settlement in a building. 

The limit of allowable settlement may be better expressed in terms of angular distortion, θ is 

θ =δ / L (4.11) 

4-10 March 2009


Where δ = differential settlement in the structure 

L = horizontal distance between the 2 points where δ is considered. 

Skelton and McDonald established that for no architectural damage, θ must be less than 1/300 for 

buildings on individual footings. As a guide, reader can refer to Table 4.3 for the typical tolerable 

values of differential settlement. 

Table 4.2 Typical Values of Tolerable Differential Settlement 

Span 

Structure 

ß 

/3 

Type of Structure 

Settlement 

profile 

Circular steel petrol or fluid 

storage tanks: 

Fixed top 

Floating top 

Tolerable differential 

settlement, ß (radians) 

0.008 

0.002 – 0.003 

Tracks for overhead 

travelling crane. 0.003 

Rigid circular ring or mat 

footing for stacks, silos, 

water tanks etc. 

Jointed rigid concrete 

pressure pipe. 

One- or two-storey steel 

framed warehouse with 

truss roof and flexible 

cladding. 

One- or two-storey houses 

or similar buildings with 

brick load-bearings walls. 

Structures with sensitive 

interior finishes such as 

plaster, ornamental stone 

or tiles. 

Multi-storey heavy concrete 

rigid framed structures on 

thick structural raft 

foundations. 

(Source: Carter M, [7]) 

0.002 

0.015 

0.006 – 0.008 

0.002 – 0.003 

0.001 -0.002 

0.0015 

Differential 

settlement 

Comments 

For floating top, value depends on 

details of top. Values apply to tanks on 

a flexible base. With rigid base slabs, 

such settlement will cause cracking and 

local buckling. 

Value taken longitudinally along track. 

Settlement between between tracks is 

not usually the controlling factor. 

Value is allowable angle change at joint. 

This is usually 2-4 times average slope 

of settlement profile. Damage to joint 

also depends on Longitudinal extension. 

Overhead crane, pipes, machinery or 

vehicles may limit tolerable values to 

less than this. 

Larger value is tolerable if most 

settlement has taken place before 

finishes are completed. 

Damage to interior or exterior finish 

may limit value. 

March 2009 4-11


4.7 PLATE LOADING TEST FOR SETTLEMENT ESTIMATION 

Guidelines and procedures for conducting plate loading tests are given in BS EN 1997-1:2004 (BSI, 

2004) and DD ENV 1997-3:2000 (BSI, 2000b). The test should mainly be used to derive 

geotechnical parameters for predicting the settlement of a shallow foundation, such as the 

deformation modulus of soil. It may be necessary to carry out a series of tests at different levels. 

The plate loading test may also be used to determine the bearing capacity of the foundation in finegrained 

soils, which is independent of the footing size. The elastic soil modulus can be determined 

using the following equation (BSI, 2000b): 

E s = q net b 1-v s 2 

δ p 

I s (4.12) 

where 

q net 

= net ground bearing pressure 

δ p 

= settlement of the test plate 

I s 

= shape factor 

b = width of the test plate 

ν s 

= Poisson’s ratio of the soil 

E s 

= Young's modulus of soil 

The method for extrapolating plate loading test results to estimate the settlement of a full-size 

footing on granular soils is not standardised. The method proposed by Terzaghi & Peck (1917) 

suggested the following approximate relationship in estimating the settlement for a full-size footing: 

δ f = δ p 2B f 

B f +b 2 (4.13) 

where: 

δ p = settlement of a 30mm square test plate 

δ f = settlement of foundation carrying the same bearing pressure 

B f = width of the shallow foundation 

B = width of the test plate 

However, the method implies that the ratio of settlement of a shallow foundation to that of a test 

plate will not be greater than 4 for any size of shallow foundation and this could under estimate the 

foundation settlement. Bjerrum & Eggestad (1913) compared the results of plate loading tests with 

settlement observed in shallow foundations. They noted that the measured foundation settlement 

was much greater than that estimated from the method of Terzaghi & Pack (1917). Terzaghi et al 

(1991) also commented that the method is unreliable and is now recognized to be an unacceptable 

simplification of the complex phenomena. 

4.8 SETTLEMENT OF RAFT/MAT FOUNDATIONS 

A raft/mat foundation is usually continuous in two directions and covers an area equal to or greater 

than the base area of the structure. A raft foundation is suitable when the underlying soils have a 

low bearing capacity or large differential settlements are anticipated. It is also suitable for ground 

containing pockets of loose and soft soils. In some instances, the raft foundation is designed as a 

cellular structure where deep hollow boxes are formed in the concrete slab. The advantage of a 

cellular raft is that it can reduce the overall weight of the foundation and consequently the net 

applied pressure on the ground. A cellular raft should be provided with sufficient stiffness to reduce 

differential settlement. 

4-12 March 2009

Chapter 


Raft foundations are relatively 

large in size. Hence, 

the bearing capacity is generally not the 

controlling factor in 

design. Differential and 

total settlements usually 

govern the 

design. A common 

approach for estimating the settlement of 

a raft foundation is to 

model the 

ground support as 

springss using the subgrade reaction method. This method suffers from a number of drawbacks. 

Firstly, the modulus 

of subgrade reaction is 

not an intrinsic soil property. It depends upon not only 

the stiffness of the soil, but also 

the dimensions of the foundation. 

Secondly, there is no interaction 

between the springs. They are 

assumed to 

be independent of each other and can only respond in 

the direction of the 

loads. BSI (2004) cautions that the subgrade 

reaction model is generally not 

appropriate for estimating the total and differential settlement of a raft foundation. Finite element 

analysiss or elastic continuum method is preferred for the design of 

raft foundations (French, 1999; 

Poulos, 

2000). 

Figure 4.9 Common Types of Raft Foundation 

March 2009 

4-13


REFERENCES 




New York, 1992, 1004 p. 




[5] Buisman, A.S.K. Results of long duration settlement tests. Proceedings of the First 

International Conference on Soil Mechanics and Foundation Engineering, Cambridge, Massachusetts, 

vol. 1, pp 103-101, 1931. 

[6] Burland, J.B. & Burbidge, M.C. Settlement of foundations on sand and gravel. Proceedings of 

Institution of Civil Engineers, Part 1, vol. 78, pp 1325-1381, 1985 

[7] Carter M. & Symons, M.V., Site Investigations and Foundations Explained, Pentech Press, 

London 

[8] CGS, Canadian Foundation Engineering Manual, (Third edition). Canadian Geotechnical 





1982, "Soil Mechanics" 


Foundations and Earth Structures 

[12] Duncan, J.M. & Poulos, H.G. (1981). Modern techniques for the analysis of engineering 

problems in soft clay. Soft Clay Engineering, Elsevier, New York, pp 317-414. 

[13] DID Malaysia, Geotechnical Guidelines for D.I.D. works 

[14] EM 1110-2-1913. Design and Construction of Levees, U.S. Army Corp of Engineer, 

Washington, DC. 

[15] French, S.E. (1999). Design of Shallow Foundations, American Society for Civil Engineers 

Press, 374 p. 

[16] Foott R. and Ladd C.C., Undrained Settlement of Plastic and Organic Clays, Journal of 

Geotechnical Engineering Division, ASCE, Vol.107, No. GT8, August 1981. 

[17] ISE (1989). Soil-structure Interaction: The Real Behaviour of Structures. The Institution of 

Structural Engineers, London, 120 p. 

[18] Koerner R.M ., Construction and Geotechnical Method in Foundation Engineering, McGraw 

Hill, 1985. 

4-14 March 2009


[19] Donald P.Coduto, Foundation Design, Principles and Practices 

[20] Ladd C.C., Foott R., Ishihara K., Schlosser F., and Roulos H.G., Stress Deformation and 

Strength Characteristics, State of the Art Report, Session I, IX ICSMFE, Tokyo, Vol. 2, 1971, pp. 421 

- 494. 


[22] Liao S.S.C. and Whitman R. V., Overburden Correction Factors for SPT' in Sand, Journal of 

the Geotechnical Engineering Division, ASCE. Vol. 112 No. 3, March 1986, pp. 373 - 377. 


[24] Mesri G., discussion of New Design Procedure for stability of Soft Clays, by Charles C. Ladd 

and Roger Foott, Journal of the Geotechnical Engineering Division, ASCE, Vol.101, No. GT4. Froc. 

Paper 10664. April 1975. pp. 409 - 412. 

[25] Mesri, G., Lo, D.O.K. & Feng, T.W. (1994). Settlement of embankments on soft clays. 

Geotechnical Special Publication 40, American Society of Civil Engineers, vol. 1, pp 8-51. 


[27] Parry, R.G. H. (1972). A direct method of estimating settlement in sands from SPT values. 

Proceedings of the Symposium on Interaction of Structures and Foundations, Midland Soil Mechanics 

and Foundation Engineering Society, Birmingham, pp 29-37. 


1974. 

[29] Poulos, H.G. & Davis, E.H. (1974). Elastic Solutions for Soil and Rock Mechanics. John Wiley 

& Sons, New York, 411 p. 

[30] Poulos, H.G. (2000). Foundation Settlement Analysis – Practice versus Research. The Eighth 

Spencer J Buchanan Lecture, Texas, 34 p. 

[31] Poulos, H.G., Carter, J.P. & Small, J.C. (2002). Foundations and retaining structures – 

research and practice. Proceedings of the Fifteenth International Conference on Soil Mechanics and 

Foundation Engineering, Istanbul, vol. 4, pp 2527-2101. 

[32] Price, G. & Wardle, I.F. (1983). Recent developments in pile/soil instrumentation systems. 

Proceedings of the International Symposium on Field Measurements in Geomechanics, Zurich, vol. 1, 

pp 2.13-2.72. 

[33] Research and practice. Proceedings of the Fifteenth International Conference on Soil 

Mechanics and Foundation Engineering, Istanbul, vol. 4, pp 2527-2101. 

[34] Skempton A.W. and D.H. McDonald, "The Allowable Settlement of Buildings", Proc. Inst. Civil 

Eng., Vo1.5 Pt.3. 1956, pp. 727-784. 

[35] Skempton A.W., "The Bearing Capacity of Clays", Building Res. Congress, London Inst. Civ. 

Engrs., div.I:180, 1951. 

[36] Smith C.N., "Soil Mechanics for Civil and Mining Engineers". 

March 2009 4-15


[37] Teng W.C., "Foundation Design", Prentice Hall, 1984. 



[39] Thompson D.M. and Shuttler R.M., "Design of riprap slope protection against wind waves", 

Report 61, London, Construction Industry Research & Information Association. 

[40] Terzaghi, K. (1955). Evaluation of coefficients of subgrade reaction. Géotechnique, vol. 5, pp 

297-321. 

[41] Tomlinson, M.J. (1994). Pile Design and Construction Practice. (Fourth edition). Spon, 411 p. 

[42] United Bureau States Department of the Interior, "Design of Small Dams” Bureau of 

Reclamation, Oxford and IBH Publishing Co., 1974. 

[43] Vesic, A.S. (1975). Bearing capacity of shallow foundations. Foundation Engineering 

Handbook, edited by Winterkorn, H.F. & Fang, H.Y., Van Nostrand Reinhold, New York, pp 121-147. 

[44] Zanen A., "Revetments", International Institute for Hydraulic and Environmental Engineering, 

Delft, Netherlands, 1978 

4-16 March 2009

CHAPTER 5 BEARING CAPACITY THEORY

Chapter 5 BEARING CAPACITY THEORY 





5.1 SHALLOW FOUNDATION ............................................................................................. 5-1 

5.1.1 Bearing Capacity of Shallow Foundation ......................................................... 5-1 

5.1.1.1 General ........................................................................................ 5-1 

5.1.1.2 General Equation For Bearing Capacity ............................................ 5-2 

5.1.2 Factors of Safety .......................................................................................... 5-5 

5.1.3 Effects of Groundwater ................................................................................. 5-5 

5.1.4 Foundation Near Crest of Slope ..................................................................... 5-6 

REFERENCES ......................................................................................................................... 5-8 

March 2009 5-i




5.1 Bearing Capacity Factors for Computing Ultimate Bearing Capacity of Shallow 

Foundations 5-4 



5.1 Generalized Loading and Geometric Parameter for a Spread Shallow Foundation 5-3 

5.2 Groundwater Cases for Bearing Capacity Analysis 5-6 

5.3 Linear Interpolation Procedures for Determining Ultimate Bearing Capacity of a 

Spread Shallow Foundation near the Crest of a Slope 5-7 

5-ii March 2009


5.1 SHALLOW FOUNDATION 

5 BEARING CAPACITY THEORY 

Shallow foundations, are generally more economical than deep foundations if they do not have to 

be installed deep into the ground and extensive ground improvement works are not required. They 

are often used to support structures at sites where ground are sufficiently strong. Unless a shallow 

foundation can be founded on strong rock, some noticeable settlement will occur. Design of 

shallow foundations should ensure that there is an adequate factor of safety against bearing failure 

of the ground, and that the settlements, including total and differential settlement, are limited to 

allowable values. 

For shallow foundations founded on granular soils, the allowable load is usually dictated by the 

allowable settlement, except where the ultimate bearing capacity is significantly affected by 

geological or geometric features. Examples of adverse geological and geometrical features are 

weak seams and sloping ground respectively. For shallow foundations founded on fine-grained soils, 

both the ultimate bearing capacity and settlements are important design considerations. 

High-rise structures or the presence of weak ground bearing materials do not necessarily stopping 

the design engineer from adopting shallow foundation system. Suitable design provision or ground 

improvement could be considered to overcome the difficulties. Some examples are given below: 

a. Design the foundations, structures and building services to accommodate the expected 

differential and total settlements. 

b. Excavate weak materials and replace them with compacted fill materials. 

c. Carry out in-situ ground improvement works to improve the properties of the bearing materials. 

Some of these methods are discussed in Chapter 9. 

d. Adopt specially designed shallow foundations, such as compensated rafts, to limit the net 

foundation loads or reduce differential settlement. 

5.1.1 Bearing Capacity of Shallow Foundation 

5.1.1.1 General 

There are a many of methods for determining the bearing capacity of shallow foundations on soils. 

A preliminary estimate of allowable bearing pressure may be obtained on the basis of soil 

descriptions. Other methods include correlating bearing pressures with results of in-situ field tests, 

such as SPT N value and tip resistance of CPT. For example, Terzaghi & Peck (1917) proposed 

allowable bearing pressure of 10 N (kPa) and 5N (kPa) for non-cohesive soils in dry and submerged 

conditions respectively. This was based on limiting the settlement of footings of up to about 1 m 

wide to less than 25 mm, even if it is founded on soils with compressible sand pockets. 

Methods based on engineering principles can be used to compute the bearing capacity of soils and 

estimate the foundation settlement. This would require carrying out adequate ground investigation 

to characterize the site, obtaining samples for laboratory tests to obtain parameters and establishing 

a reliable model. Designs following this approach normally result in bearing pressures higher than 

the presumed allowable bearing pressures given in codes of practice. 

March 2009 5-1


5.1.1.2 General Equation For Bearing Capacity 

Various equations have been established for calculating the bearing of shallow foundation. A 

comprehensive one which takes into consideration the shape of the foundation, inclination of 

loading, the base of the foundation and ground surface is as follows 

(GEO, 1993): 

q u = Q u 

Bf'Lf' 

c'Nc ζ cs 

ζ ci 

ζ ct 

ζ cg 

+ 0.5 Bf' γs' Nγ ζ γs 

ζ γi 

ζ γt 

ζ γg 

+ q Nq ζ qs 

ζ qi 

ζ qt 

ζ qg 

(5.1) 

Where: 

Nc, Nγ, Nq = general bearing capacity factors which determine the capacity of a long strip 

footing acting on the surface of a soil in a homogenous half space 

Q u = ultimate resistance against bearing capacity failure 

q u = ultimate bearing capacity of foundation 

q 

= overburden pressure at the level of foundation base 

c’ = effective cohesion of soil 

γs’ = effective unit weight of the soil 

Bf = least dimension of footing 

Lf 

= longer dimension of footing 

Bf’ = Bf – 2e B 

Lf’ = Lf – 2e L 

e L = eccentricity of load along L direction 

e B = eccentricity of load along B direction 

ζ cs 

, ζ γs 

, ζ qs 

= influence factors for shape of shallow foundation 

ζ ci 

, ζ γi 

, ζ qi 

= influence factors for inclination road 

ζ cg 

, ζ γg 

, ζ qg 

= influence factors for ground surface 

ζ ct 

, ζ γt 

, ζ qt 

= influence factors for tilting of foundation base 

Figure 5.1 shows the generalized loading and geometric parameters for the design of a shallow 

foundation. The bearing capacity factors are given in Table 5.1. Equation 5.1 is applicable for the 

general shear type of failure of a shallow foundation, which is founded at a depth less than the 

foundation width. This failure mode is applicable to soils that are not highly compressible and have a 

certain shear strength, e.g. in dense sand. If the soils are highly compressible, e.g: in loose sands, 

punching failure may occur. Vesic (1975) recommended using a rigidity index of soil to define 

whether punching failure is likely to occur. In such case, the ultimate bearing capacity of the 

foundation can be evaluated based on Equation 5.1 with an additional set of influence factors for soil 

compressibility (Vesic,1975). 

5-2 March 2009


Figure 5.4 Generalized Loading and Geometric Parameter for a Spread Shallow Foundation 

March 2009 

5-3


Table 5.1 Bearing Capacity Factors for Computing Ultimate Bearing Capacity of Shallow Foundations 

5-4 

March 2009


5.1.2 Factors of Safety 

The net allowable bearing pressure of a shallow foundation resting on soils is obtained by applying 

a factor of safety to the net ultimate bearing capacity i.e. 

q u = q ult 

F 

(5.2) 

where 

q ult = ultimate net bearing capacity 

q u = allowable bearing capacity 

F = Factor of safety 

The net ultimate bearing capacity should be taken as (q u – γ D f ) where D f is the depth of soil 

above the base of the foundation and γ is the bulk unit weight of the soil. The selection of the 

appropriate factor of safety should consider factors such as: 

(a) The frequency and likelihood of the applied loads (including different combination of dead 

load and live loads) reaching the maximum design level. 

(b) Soil variability, e.g. soil profiles and shear strength parameters. The ground investigation 

helps increase the reliability of the site characterization. 

(c) The importance of the structures and the consequences of their failures. 

In general, the minimum required factor of safety against bearing failure of a shallow foundation is 

in the range of 2.5 to 3.5. For most applications, a minimum factor of safety of 3.0 is adequate. 

Although the factor of safety is applied to the bearing capacity at failure, it is frequently used to 

limit the settlement of the foundation. 

5.1.3 Effects of Groundwater 

The ultimate bearing capacity depends on the effective unit weight of the soil. Where groundwater 

is present, the effective stress and shear strength along failure plane will be smaller and the bearing 

capacity will be reduced. The effect of groundwater is accounted for by adjusting the γ s ' in equation 

5.1. and the three possible cases as shown in Figure 5.2 and describe below: 

a) Case 1: D w < D 

Use γ’ = γ b = γ - γ w 

where γ b = weighted average buoyant unit weight 

b) Case 2: D < D w < D + B 

Use ′ w 

1- Dw-D 

B 

c) Case 3: D + B < D w (no groundwater correction is necessary ) 

Use γ’ = γ 

March 2009 5-5


D w 

D w 

D 

D w 

D + B 

Lower Limit of Zone of influence 

Case 1 Case 2 

Case 3 

Figure 5.5 Groundwater Cases for Bearing Capacity Analysis 

5.1.4 Foundation Near Crest of Slope 

An approximate method is given in Geoguide 1: Guide to Retaining Wall Design (GEO HONG KONG, 

1993) to determine the ultimate bearing capacity of a foundation near the crest of a slope. The 

ultimate bearing capacity can be obtained by linear interpolation between the value for the 

foundation resting at the edge of the slope and that at a distance of four times the foundation 

width from the crest. Equation 2.2 in section 2.2 can be used to estimate the ultimate bearing 

capacity for the foundation resting on the slope crest. Figure 5.3 summarises the procedures for 

the linear interpolation. 

5-6 March 2009


Figure 

5.6 Linear Interpolationn Procedures 

for Determining Ultimate Bearing Capacity of a Spread 

Shallow Foundation near the 

Crest of a Slope 

March 2009 

5-7


REFERENCES 

[1] Bishop A.V and Henkel D.J., The Measurement of Soil Properties in the Triaxial Test, E.Arnold, 

1962. 

[2] Bowles, J.E. Foundation Analysis and Design. (Fourth edition). McGraw-Hill International, New 

York, 1992, 1004 p. 




[5] Buisman, A.S.K. Results of long duration settlement tests, Proceedings of the First 


vol. 1, pp 103-101, 1931. 


London 

[7] CGS, Canadian Foundation Engineering Manual, (Third edition). Canadian Geotechnical Society, 

Ottawa, 1992, 512 p. 

[8] Das, B.M., Principles of Geotechnical Engineering, PWK-Kent Publishing Company , Boston,MA., 

1990 


[10] Dept. of the Navy, Bureau of Yards and Docks, Washington D.C., NAVFAC DM-7.1, May 1982, 

Soil Mechanics 






Press, 374 p. 

[14] GCO (1990) Review of Design Method for Excavation, Geotechnical Control Office, Hong Kong 

[15] Hansen J.B . A Revised and Extended Formula for Bearing Capacity, Danish Geotechnical 

Institute, Bulletin No. 28; October 1968. 

[16] Holtz, R.D., Kovacs, W.D. An Introduction to Geotechnical Engineering, Prentice-Hall, Inc. New 

Jersey 



[18] Ladd C.C., Foott R., Ishihara K., Schlosser F., and Roulos H.G., "Stress Deformation and 

Strength Characteristics", State of the Art Report, Session I, IX ICSMFE, Tokyo, Vol. 2, 1971, pp. 421 

- 494. 

5-8 March 2009



[20] Liao S.S.C. and Whitman R. V., Overburden Correction Factors for SPI' in Sand, Journal of the 

Geotechnical Engineering Division, ASCE. Vol. 112 No. 3, March 1986, pp. 373 - 377. 




1974. 

[24] Poulos, H.G. & Davis, E.H. (1974). Elastic Solutions for Soil and Rock Mechanics. John Wiley & 

Sons, New York, 411 p. 

[25] Poulos, H.G., Carter, J.P. & Small, J.C. (2002). Foundations and retaining structures – research 

and practice. Proceedings of the Fifteenth International Conference on Soil Mechanics and 




[27] Skempton A.W., The Bearing Capacity of Clays, Building Res. Congress, London Inst. Civ. 

Engrs., div.I:180, 1951. 






297-321. 





March 2009 5-9



5-10 March 2009

CHAPTER 6 SLOPE STABILITY

Chapter 6 SLOPE STABILITY 


Table of Contents .................................................................................................................... 6-I 

List of Tables ....................................................................................................................... 6-III 

List of Figures ...................................................................................................................... 6-III 

6.1 INTRODUCTION ..................................................................................................... 6-1 

6.2 TYPE OF SLOPE INSTABILITIES ............................................................................... 6-1 

6.2.1 Infinite Slope Failure .............................................................................. 6-1 

6.2.2 Sliding Block Failure ............................................................................... 6-1 

6.2.3 Circular Arc Failure ................................................................................. 6-2 

6.3 GENERAL PROCEDURE FOR ANALYSIS ..................................................................... 6-3 

6.3.1 Obtaining Subsurface Information ........................................................... 6-3 

6.3.2 Determining of Soil Shear Strengths ........................................................ 6-3 

6.3.3 Determining a Potential Slide Failure Surface ............................................ 6-3 

6.4 PRINCIPLES OF ANALYSIS ...................................................................................... 6-4 

6.4.1 Method of Analysis ................................................................................. 6-4 

6.4.2 Stages of Stress Analysis ........................................................................ 6-4 

6.4.2.1 Short-Term (or At-the-end-of-construction) .............................. 6-4 

6.4.2.2 Long-term ............................................................................. 6-5 

6.5 CIRCULAR ARC ANALYSIS ....................................................................................... 6-5 

6.5.1 General Principles................................................................................... 6-5 

6.5.2 Location of the Critical Slip Surface .......................................................... 6-6 

6.5.4 Required Safety Factors .......................................................................... 6-7 

6.5.5 Cut Slope in Clay .................................................................................... 6-7 

6.5.6 Filled Slope/Embankment on Clay ............................................................ 6-8 

6.5.7 Effects of Water ..................................................................................... 6-8 

6.5.7.1 Effects on Cohesionless Soils ................................................... 6-9 

6.5.7.2 Effects on Cohesive Soils ........................................................ 6-9 

6.5.8 Method of Slides for Circular Failure ......................................................... 6-9 

6.5.9 Finite Element Methods ........................................................................ 6-11 

6.6 SLIDING BLOCK FAILURE ...................................................................................... 6-12 

6.7 SLOPE STABILIZATION METHODS ......................................................................... 6-13 

6.7.1 Slope Flattening ................................................................................... 6-13 

6.7.2 Drainage ............................................................................................. 6-13 

6.7.3 Buttressing or Counter Berm ................................................................. 6-14 

6.7.4 Soil Nailing .......................................................................................... 6-14 

March 2009 6-i


6.7.5 Geo-Synthetically Reinforcements .......................................................... 6-15 

6.7.6 Retaining Walls .................................................................................... 6-15 

REFERENCES ....................................................................................................................... 6-16 

APPENDIX 6.A WORKED EXAMPLE: SLOPE STABILITY .................................................. 6A-1 

6-ii March 2009




6.1 Undrained Shear Strength and Consistency of Cohesive Soils (After Terzaghi & Peck 

and ASTM D2488-90) 6-5 

6.2 Typical Drained Parameters For Effective Stress Analysis 6-5 

6.3 Recommended Factors Of Safety 6-7 

6.4 Guideline to Selection of Method of Slope Stability Analysis (After FHWA, Soils and 

Foundation Reference Manual) 6-11 

6.5 Summary of Results 6A-2 



6.1 Infinite Slope Failure 6-1 

6.2 Sliding Block Failure Mechanism 6-2 

6.3 Example of Circular Arc Failure Mechanism 6-2 

6.4 Typical Circular Arc Failure Mechanism 6-6 

6.5 Relationship Of Total Stress, Pore Pressure And Time 6-8 

6.6 Effects Of Water Content On Cohesive Strength 6-9 

6.7 Method of Slides 6-10 

6.8 Geometric And Force Components For Sliding Block Analysis 6-12 

6.9 Schematic View of Slope Regrading Work 6-13 

6.10 Good Drainage System Critical to Stability of Slope 6-14 

6.11 Butresses or Counter Berm for Slope Stabilsation 6-14 

6.12 Typical Details of Soil Nail 6-15 

6.13 Related Slope Configuration 6A-1 

6.14 Stability Analysis of an Embankment Uses SLOPE/W Software 6A-3 

March 2009 

6-iii



6-iv March 2009


6 SLOPE STABILITY 

6.1 

INTRODUCTION 

Slope stability addresses the tendency of soil masses to attain an 

equilibrium 

state between the 

strength of the soil and the force of gravity. In JPS, slope stability problems most often occur in the 

construction of embankment over soft soils, and the instability of waterway slope (e.g. river and 

pond) due to seepage, drawdown, or erosion by flowing water. Placement of stockpiles, heavy 

equipment, or other surcharges may also cause instabilities of 

the slope, particularly 

during 

construction stage. In general, 

altered slope, whether man-made or natural need to be analyzed 

and checked to ensure that it has adequate factor of safety against slope failure. 

The factor of safety 

against slope failure is defined as the ratio of the resisting forces to the 

driving 

forces tending to cause movement for a given failure configuration. 

The analysiss of slope stability is 

therefore the analytical procedure of determining the most critical, i.e. the lowest factor of safety of 

a given 

or proposed 

slope configuration. 

6.2 

TYPE OF SLOPE INSTABILITIES 

In general, slope stability problems commonly encountered in JPS can be categories into three 

types, namely: 

6.2.1 

Infinite Slope Failure 

A slope 

that extends for a relatively long distance and has a consistent subsurface profile may be 

analyzed as an infinite slope, see Figure 6.1. The failure 

plane for this case is parallel to the surface 

of the slope and the 

limit equilibrium method can be applied readily. 

Figure 6.1 Infinite Slope Failure 

6.2.2 

Sliding Block Failure 

Sliding block failure 

occurs when the wedgee type of sliding mass that cut through the fill and a thin 

layer of 

weak soil essentially moves as a block. This concept is as shown in Figure 6.2. 

March 2009 

6-1


Fill 

Fill 

Firm 

soil 

Sliding 

Thin Seam of 

Weak 

Clay 

Material of 

General Low 

Permeability 

Sliding 

Lens of 

Sand 

without Friction 

Fill 

Sliding 

Shallow Layer of Weak Soil 

Firm 

Soil 

Figure 6.2 Sliding Block Failure Mechanism 

6.2.3 

Circular Arc Failure 

All of the limit equilibrium methods requiree that a potential slip surface to be assumed in order to 

calculate the factor 

of safety. For computational simplicity the slip 

surface is often assumed to be 

circular, particularly 

for relatively homogeneous soil condition. Calculations are repeated for a 

sufficient number of trial slip surfaces to 

ensure that the minimum factor of safety has been 

obtained. 

Circularr arc failure occurs when 

the ground sink down and the adjacent ground rises and the failure 

surface 

follows a circular arc as 

illustrated in Figure 6.3. This type 

of failure shall be discussed in 

more detail in this chapter as it is a very 

common mode of failure especially 

in river bank and 

embankment in soft ground. 

Figure 6.3 Example of Circular Arc Failure Mechanism 

6-2 

March 2009


6.3 GENERAL PROCEDURE FOR ANALYSIS 

In general, analysis of slope stability would involves three basic parts: 

a) Obtaining subsurface information 

b) Determining appropriate soil shear strengths and 

c) Determining a potential slide failure surface which provides the minimum safety factor 

against failure under the various conditions 

6.3.1 Obtaining Subsurface Information 

Previous works carried out at the site of interest generally can provide some subsurface information 

which are usually indicated in the design report or construction plans. The bore logs obtained may 

or may not be located close to the site and the engineer must determine if additional subsurface 

information is required. Additional boring(s) at the site are generally preferable. Other completed 

work in the nearby vicinity may also provide useful information. Soil type, thickness of each soil 

zone, depth to bedrock, and groundwater conditions must be known to proceed with a slope 

stability analysis. Reader can refer to Volume 6 Part 2 for further information on this matter. 

Before any analysis being carried out, it is always advisable to carry out geomorphological mapping 

of the project area. The observations during the mapping works can sometimes help significantly in 

deciding the types of tests, site investigation works and strengthening measures. The tell tale signs 

observed during the mapping works i.e., water seepages, ground saturation, erosion; mode of 

failure (deep seated or shallow slip) can be the references in the analysis and design stage. These 

geomorphologic features are always tie up with the estimation of the design parameter i.e., ground 

water condition, drainage adequacy and inherent properties (existence of discontinuities) which are 

difficult to retrieve from site investigation works. 

6.3.2 Determining of Soil Shear Strengths 

The shear strength parameters of the embankment soil are normally defined in terms of a friction 

component (φ ) and a cohesion component (c). Shear strengths are usually determined from 

laboratory tests performed on specimens prepared by compaction in the laboratory or undisturbed 

samples obtained from exploratory soil borings. The laboratory test data may be supplemented 

with in situ field tests and correlations between shear strength parameters and other soil properties 

such as grain size, plasticity, and Standard Penetration Resistance (N) values. For a more detail 

discussion, reader can refer to Item. 3.3 of this Part. 

In general, for drained shear parameters for effective stress analysis, consolidated undrained (CU) 

can be used to obtained the effective soil strength parameter i.e., effective frictional angle φ‘ and 

effective cohesion c’. Shear box test can also be used in determining the strength parameter. The 

shear box sample shall be soaked in water for saturation and the shear rate shall be low to avoid 

misleading results. High cohesion (sometimes as high as 10kPa) and low frictional angle are the 

common error obtained from such tests if the saturation procedure is omitted. 

6.3.3 Determining a Potential Slide Failure Surface 

All of the limit equilibrium methods require that a potential slip surface to be assumed in order to 

calculate the factor of safety. Circular slip surfaces can be assumed if the soil conditions are 

revealed to be relatively homogeneous. If the soil conditions are not homogeneous or if geologic 

anomalies appear, slope failures may occur on non-circular slip surfaces. The shape of the failure 

surface will depend on the problem geometry and stratigraphy, material characteristics (especially 

anisotropy), and the capabilities of the analysis procedure used. Commercially available computer 

March 2009 6-3


programs such as SLOPE/W and STABL, which offer several analysis procedures, are useful for 

slope stability assessment. 

6.4 PRINCIPLES OF ANALYSIS 

6.4.1 Method of Analysis 

The methods for analysis of slope stability broadly used in engineering practice are limit equilibrium 

methods and finite element methods. The limit equilibrium method of slope stability analysis is 

used to evaluate the equilibrium of a soil mass tending to move down slope under the influence of 

gravity. A comparison is made between forces, moments, or stresses tending to cause instability of 

the mass, and those that resist instability. Two-dimensional (2-D) sections are analyzed and plane 

strain conditions are assumed. These methods assume that the shear strengths of the materials 

along the potential failure surface are governed by linear (Mohr-Coulomb) or nonlinear relationships 

between shear strength and the normal stress on the failure surface. 

Where estimates of movements as well as factor of safety are required to achieve design 

objectives, the effort required to perform finite element analysis can be justified. However, finite 

element analysis requires considerably more time and effort, compared to the limit equilibrium 

analysis and additional data related to stress-strain behavior of materials. Therefore, the use of 

finite element analysis is not justified for the sole purpose of calculating factors of safety. 

6.4.2 Stages of Stress Analysis 

As mentioned in Para 3.3, shear strength of the soil varies with time. Thus, in slope stability 

analysis, it is important for the designer to understand and determine at which point in time i.e. 

before, during or after construction that is more critical and yield the lowest factor of safety. 

Generally, the two conditions considered are: 

6.4.2.1 Short-Term (or At-the-end-of-construction) 

Analyses of the short-term condition of stability are normally performed in terms of total stress 

(using undrained shear strength parameters), with the assumption that any pore water pressure set 

up by the construction activity will not dissipate at all. However, in some construction works such as 

large earth dams or embankments, the construction period is relatively long, and some dissipation 

of the excess pore water pressure is likely. Under these conditions, a total stress analysis would 

yield a value of factor of safety on the low side, possibly resulting in un-economic design. 

For undrained shear strength of saturated soil, φ can be assumed as zero and knowledge of the 

pore water pressure (i.e. the phreatic line) is not necessary since total stress can be expressed 

independently of effective stress at failure. For instance, the total stress analysis must be used for 

the construction of coastal bund in soft clay and it usually gives the worst critical factor of safety. 

Unconsolidated Undrained (UU) Triaxial test is usually used to obtain the undrained strength 

parameter of the soil. Extra care shall be given during the test when the soil samples are not fully 

saturated. For soft to very soft clay such as coastal alluvium clay, in-situ strength test using in-situ 

vane shear test should be used to determine the undrained shear strength. Typical values of 

undrained shear strength for Malaysia coastal alluvium clay ranges from 10 to 20 kPa. 

Table 6.1 gives some typical values of undrained shear strength, c which may be used for 

preliminary analysis or to check laboratory test results 

6-4 March 2009


Table 6.1 Undrained Shear Strength and Consistency of Cohesive Soils 

Consistency 

Undrained Shear 

Strength, S u (kPa) 

Visual Identification 

Very soft < 12 Thumb can penetrate more than 25 mm 

Soft 12 – 25 Thumb can penetrate about 25 mm 

Medium 25 -50 

Thumb can penetrate with moderate 

effort 

Stiff 50 – 100 Thumb will indent soil about 8 mm 

Very stiff 100 – 200 

Thumb will not indent but readily indent 

with thumbnail 

(After Terzaghi & Peck and ASTM D2488-90) 

6.4.2.2 Long-term 

Long-term stability analysis is normally carried out using effective stress analysis with drained shear 

strength parameters. For cohesive or clayey soil, total stress analysis (for short-term) in addition to 

the effective stress analysis (for long-term) are carried out to determine the most critical factor of 

safety. As granular or sandy soils are more permeable than cohesive or clayey soils, drainage of 

excess pore pressure in sandy soil occurs much more rapidly. Hence, only effective stress analysis is 

usually required. 

Effective stress analysis requires the estimation of the drained strength parameters c’, φ’ and pore 

pressures. For pure free draining sands, φ = φ’ and c = 0. Under conditions of steady seepage, the 

phreatic line can be obtained from the flow net. 

Some common drained strength parameters, φ' and c’ adopted in the slope analysis are as follows:- 

Table 6.2 Typical Drained Parameters For Effective Stress Analysis 

Soil type Effective friction angle φ‘ Effective cohesion c’ 

Well compacted soil 28 o – 30 o 2 – 5 kPa 

Residual soil grade V to VI 30 o – 32 o 5 – 10kPa 

Residual soil grade IV to V 32 o – 35 o 10 – 15kPa 

Note:- 

• The values above are just for references. Test shall be carried out before any 

analysis is carried out. It is advisable to limit the cohesion to not more than 

15kPa even with lab test results. The cohesion shows in test are sometimes 

apparent and the changes are subjected to external factors i.e., weathering 

process etc 

• Description of grade of residual soil: 

Grade VI = residual soil : Grade V = completely weathered rock ; Grade IV = 

highly weathered 

6.5 CIRCULAR ARC ANALYSIS 

6.5.1 General Principles 

Figure 6.4 shows a potential slide mass defined by a predetermined circular arc slip surface. If the 

shear resistance of the soil along the slip surface exceeds that necessary to provide equilibrium, the 

mass is stable. If the shear resistance is insufficient, the mass is unstable. Thus, the stability or 

instability of the mass depends on its weight, the external forces acting on it, the shear strengths 

March 2009 6-5


and pore-water pressures along the slip surface. 

Circular arc slip surface is often used because it simplifies the calculations by just conveniently 

summing up the moments or forces about the center of the circle. Also, circular slip surfaces are 

generally sufficient for analyzing relatively homogeneous embankments or slopes. 

Fill Surface 

after Failure 

L w 

Fill 

Weight 

Force 

Center 

L s 

Failure 

Case 

Soft Clay 

Resistance 

Force 

Sum of Shear Strength 

along Arc 

Figure 6.4 Typical Circular Arc Failure Mechanism 

The requirement for static equilibrium of the soil mass are used to compute a factor of safety with 

respect to shear strength. The factor of safety is defined as the ratio of the available shear 

resistance to the driving force that can cause movement of the slope. In Figure 6.4, the factor of 

safety (FOS) is 

Resisting Moment Total shear strength x Ls 

FOS = = 

Driving Moment Weight force × Lw 

(6.1) 

Limit equilibrium analysis assumes the factor of safety is the same along the entire slip surface. A 

value of factor of safety greater than 1.0 indicates that shear resistance exceeds the required for 

equilibrium and that the slope will be stable with respect to sliding along the assumed particular slip 

surface analyzed. A value of factor of safety less than 1.0 indicates that the slope will be unstable. 

6.5.2 Location of the Critical Slip Surface 

The critical slip surface is defined as the surface with the lowest factor of safety. Because different 

methods of analysis like Bishop’s, Janbu’s and Spencer’s adopt different assumptions, the location 

of the critical slip surface can vary among different methods of analysis. The critical slip surface for 

a given problem analyzed by a given method is found by a systematic procedure of generating trial 

slip surfaces until the one with the minimum factor of safety is obtained. Searching schemes may 

vary with the assumed shape of the slip surface and the computer program used. 

All external loadings imposed on the embankment or ground surface should be represented in slope 

stability analysis, including loads imposed by water pressures, structures, surcharge loads, anchor 

forces, or other causes. 

6-6 March 2009


6.5.4 Required Safety Factors 

Appropriate factors of safety are required to ensure adequate performance of embankments 

throughout their design lives. Two of the most important considerations that determine appropriate 

magnitudes for factor of safety are uncertainties in the conditions being analyzed, including shear 

strengths and consequences of failure (both economic loss and loss of life) or unacceptable 

performance. 

The values of factor of safety listed in Table 6.3 provide a guidance and are not prescribed for 

slopes of embankment dams. Higher or lower values might be warranted in respect of the degree of 

uncertainties in the conditions being analyzed, economic loss and loss of life. 

Type of slopes 

6.5.5 Cut Slope in Clay 

Table 6.3 Recommended Factors Of Safety 

End of construction 

(short-term) 

Long-term (steadystage 

seepage) 

Rapid 

drawdown 3 

1. Embankment and 

Natural Slope 1 1.3 1.4 1.1 – 1.2 4 

2. Cut or Excavated Slope 2 1.3 1.4 1.1 - 1.2 4 

Notes 

1. Applicable to filling for river bank, water retention facilities, levees, sea wall, stockpiles, earth 

retaining works. It also includes natural slopes such as river bank and valley slopes. 

2. Applicable to excavated slope including foundation excavation, excavated river and retention 

facilities, sea wall and other earth retaining works. 

3. Rapid drawdown occurs when it is assumed that drawdown is very fast, and no drainage 

occurs in materials with low permeability; thus the term “sudden” drawdown. 

4. For submerged or partially submerged slopes, the possibility of low water events and rapid 

drawdown should be considered. FOS of 1.1 to 1.2 for rapid drawdown recommended here are for 

cases where rapid drawdown represents an infrequent loading condition. In cases where rapid 

drawdown represents a frequent loading condition, as in river bank subjected fluctuations in water 

level and pumped storage projects, the factor of safety should be higher. 

For cut slope, the effective stress reduces with time owing to the stress relief after removal of load. 

This reduction will allow the clay to expand and absorb water, which will lead to a decrease in the 

clay strength with time. For this reason, the factor of safety of a cut slope in clay may decrease 

with time. Cut slopes in clay should be designed by using effective strength parameters and the 

effective stresses that will exist in the soil after the pore pressures have come into equilibrium 

under steady seepage condition. 

These changes in the values of total stress and pore pressure with time are shown here in Figure 

6.5(a). 

March 2009 6-7


a 

σ’ 

σ 

u 

Increase in pore pressure 

Excavation/cut 

Time 

b 

decrease in pore pressure 

σ’ 

σ 

u 

Construction/fill 

Time 

During slope cutting, frequent inspections and mapping shall be carried out by experience geologist 

to ensure no adverse “inherent” geological features i.e., soil bedding, relicts and rock discontinuities 

(if rock cutting). If these adverse features are found on slope outcrop, strengthening measures 

such as soil nailing can be specified to improve the stability of the slope. Horizontal drains can be 

installed at areas where water seepages are found during cutting to lower the ground water table. 

Always avoid cutting slope with large catchment behind the slope. Area with large catchment 

always associated with high ground water table. If it is unavoidable, Horizontal drains and deep 

trench drains shall be included in the design to lower the ground water table 

6.5.6 Filled Slope/Embankment on Clay 

Excess pore water pressures are created when fills are placed on clay or silt. Provided the applied 

loads do not cause the undrained shear strength of the clay or silt to be exceeded, as the excess 

pore water pressure dissipates consolidation will occur, and the shear strength increases with time 

as illustrated in Figure 6.5(b). For this reason, the factor of safety increases with time under the 

load of the fill. Hence, the most critical state for the stability of an filled embankment is normally 

the short-term or end-of-construction condition where total stress analysis with undrained shear 

parameters are required. 

6.5.7 Effects of Water 

Figure 6.5 Relationship Of Total Stress, Pore Pressure And Time 

Besides gravity, water (both surface and ground water) is a major factor in slope instability. In 

addition, ground water table induced failure is always deep seated and catastrophic. Ground water 

table is one of the most difficult parameter to be assumed or estimated. Hence, if necessary 

standpipes or piezometers can be installed to monitori and ascertain the fluatuation and worst 

ground water levels to be used either in design or verification of design. 

If the slope is subjected to inundation and changes in the water levels such as dam, pond, or river 

subjected to tidal effects, the designer should consider the possible effects of rapid draw down of 

water levels in the stability analysis. For rapid drawdown analysis of soils with low permeability (less 

than 10 -4 cm/sec), it is assumed that the drop in water level is so fast that no drainage can occur in 

the soil. For this prupose, drained strengths with appropriate phreatic line are used for stability 

analysis. 

6-8 March 2009


Instability of natural slopes is often related to high internal water pressures associated with wet 

weather periods. It is appropriate to analyze such conditions as long-term, steady-state seepage 

conditions, using drained strengths and the highest probable position of the piezometric surface 

within the slope. 

6.5.7.1 Effects on Cohesionless Soils 

In cohesionless soils, water does not affect the angle of internal friction (φ ’). The effect of water 

on cohesionless soils below the water table is to decrease the intergranular stress between soil 

grains (efffective normal stress, σ n '), which decreases the frictional shearing resistance. 

6.5.7.2 Effects on Cohesive Soils 

An increase in absorbed moisture is a major factor in the decrease in strength of cohesive soils as 

shown schematically in Figure 6.6. Water absorbed by clay minerals causes increased water 

contents that decrease the cohesion of clayey soils. These effects are amplified if the clay mineral 

happens to be expansive, e.g., montmorillonite. Some weak rocks such as shales, claystones, and 

siltstones tend to disintegrate into a clay soil if water is allowed to percolate into them. This 

transformation from rock to clay often leads to settlement and/or shear failure of the slope. 

cohesive strength 

water content 

Figure 6.6 Effects Of Water Content On Cohesive Strength 

6.5.8 Method of Slides for Circular Failure 

For slope stability analysis, the method of dividing the soil mass into vertical slides is most 

commonly used and illustrated in Figure 6.5 (a). The forces acting on each slide is shown in Figure 

6.7 (b) 

March 2009 6-9


(a) Method of Slides 

(b) Forces on a slide with effect of water 

Figure 6.7 Method of Slides 

Fellenius’s method of slides is one of the oldest methods used. Subsequently, several other 

methods basing on the method of slides were developed which include Bishop’s Simplified Method, 

Janbu’s Simplified Method, Morgenstern and Price’s Method and Spencer’s Method. Fellenius’s 

method is normally more conservative and gives unrealistically lower factors of safety than other 

more refined methods. The only reason this method is discussed here is to demonstrate the basic 

principles of slope stability. Reader can refer to Appendix A Example A.1 on the application of 

Fellenius’s Method of slides in deriving the factor of safety. 

Various methods may result in different values of factor of safety because: 

(a) the various methods employ different assumptions to make the problem statically 

determinate 

(b) some of the methods do not satisfy all conditions of equilibrium. 

6-10 March 2009


Table 6.4 Guideline to Selection of Method of Slope Stability Analysis (After FHWA, Soils and 

Foundation Reference Manual) 

Foundation 

Soil Type 

Cohesive 

Granular 

Type of 

Analysis 

Short-term or 

end of 

construction 

Stage 

construction 

(embankment 

s on soft clays 

– build 

embankment 

in stages with 

waiting 

periods to 

take 

advantage of 

clay strength 

gain due to 

consolidation 

Long-term 

(embankment 

on soft clays 

and clay cut 

slopes. 

Existing failure 

planes 

All types 

Source of Strength Parameters Remarks (see Note 1) 

• UU or field vane shear test 

or CU triaxial test. 

• Undrained strength 

parameters tested at p 0 

(ground overburden stress) 

• CU triaxial test. Some 

samples should be 

consolidated to higher than 

existing in-situ stress to 

determine clay strength gain 

due to consolidation under 

staged fill heights. 

• Use undrained strength 

parameters at appropriate p 0 

for staged height 

• CU triaxial test with pore 

water pressure 

measurements or CD triaxial 

test. 

• Use effective strength 

parameters. 

• Direct shear or direct simple 

shear test. Slow strain rate 

and large deflection needed. 

• Use residual strength 

parameters. 

• Obtain effective friction 

angle from charts of 

standard penetration 

resistance (SPT) versus 

friction angle or from direct 

shear tests. 

Use Bishop Method. An angle of 

internal friction should not be 

used to represent an increase of 

shear strength with depth. 

Use Bishop Method at each 

stage of embankment height. 

Consider that clay shear 

strength will increase with 

consolidation under each stage. 

Consolidation test data needed 

to estimate length of waiting 

periods between embankment 

stages. Piezometers and 

settlement devices should be 

used to monitor pore water 

pressure dissipation and 

consolidation during 

construction 

Use Bishop Method with 

combination of cohesion and 

angle of internal friction 

(effective strength parameters 

from laboratory test). 

Use Bishop, Janbu or Spencer 

Method to duplicate previous 

shear surface. 

Use Bishop Method with an 

effective stress analysis. 

Note 1: Methods recommended represent minimum requirement. More rigorous methods such as 

Spencer’s method should be used when a computer program has such capabilities. 

6.5.9 Finite Element Methods 

The finite element methods can be used to compute stresses and displacements in earth structures 

caused by applied loads. The method is particularly useful for soil-structure interaction problems, in 

which structural members interact with a soil mass. The stability of a slope cannot be determined 

directly from finite element analysis, but the computed stresses in a slope can be used to compute 

a factor of safety. Use of the finite element methods for stability problems is a complex and timeconsuming 

process. 

March 2009 6-11


Finite element analysis can provide estimates of displacements and construction pore water 

pressures. This is useful for the field control of construction works, or when there is concern for 

damage to adjacent structures. If the displacements and pore water pressures measured in the 

field differ greatly from those computed, the reason for the difference should be investigated. 

Finite element analysis provides displacement pattern which may show potential and possibly 

complex failure mechanisms. The validity of the factor of safety obtained from limit equilibrium 

analysis depends on locating the most critical potential slip surfaces. In complex conditions, it is 

often difficult to anticipate failure modes, particularly if reinforcement or structural members such 

as geotextiles, concrete retaining walls, or sheet piles are included. Once a potential failure 

mechanism is recognized, the factor of safety against a shear failure developing by that mode can 

be computed using conventional limit equilibrium procedures. 

Finite element analysis provides estimates of mobilized stresses and forces. The finite element 

method may be particularly useful in judging what strengths should be used when materials have 

very dissimilar stress-strain and strength properties, i.e., where strain compatibility is an issue. 

The finite element methods can help to identify local regions where “overstress” may occur and 

cause cracking in brittle and strain softening materials. 

6.6 SLIDING BLOCK FAILURE 

Block slide failure mechanisms are defined by dividing into straight line segments defining an active 

wedge, central block, and passive wedge. An example of the wedge is shown in Figure 6.8 

Figure 6.8 Geometric And Force Components For Sliding Block Analysis 

6-12 March 2009


The factor of safety for the wedge can be and computed by: 

FOS = 

Horizontal Resisting Forces 

Horizontal Driving forces 

= P p + cL 

P a 

(6.2) 

P a = Active force (driving) 

P p = Passive force (resisting) 

cL = Resisting force due to cohesive clay 

For method of computation of the active force and passive forces reader can refer to the Chapter 7 

on retaining wall. 

6.7 SLOPE STABILIZATION METHODS 

Slope stabilization methods generally aim to reduce driving forces, increase resisting forces, or 

both. Driving forces can be reduced by excavation of materials from appropriate part of the 

unsuitable ground and drainage of water to reduce the hydrostatic pressures acting on the unstable 

zone. Resisting forces can be increased by introducing soil reinforcements, such as soil nails and 

geo-synthetic materials, and retaining structures or other supports. 

6.7.1 Slope Flattening 

Slope flattening is a common method for increasing the stability of a slope by reducing the driving 

forces that contribute to movements. Often, it is the first option to be considered when stabilizing a 

slope. 

Existing Slope Profile 

Regrading Slope Profile 

6.7.2 Drainage 

Figure 6.9 Schematic View of Slope Regrading Work 

Surface (berm, toe, interceptor, and cascade drains) and subsurface (horizontal drains and gravel 

trenches) drainages are essential for treatment of any slide or potential slide. Proper drainage 

system can reduce the destabilizing hydrostatic and seepage forces on a slope as well as the risk of 

erosion. 

March 2009 6-13


For surface drainages, cast in-situ drains 

(both berm 

drains and cut off drains) are strongly 

recommended to avoid possible water infiltration through the poorly constructed gaps between 

precast 

drain sections. V-shape 

drain should be used due to the effect of “self cleaning” even with 

little water in the drain. 

Figure 6.10 Good Drainage System Critical to Stability of Slope 

6.7.3 

Buttressing or Counter Berm 

Buttressing is a technique used to offset or counter the driving forces of a slope by externally 

applied 

force system 

that increases the resisting forces. Buttressess may consist of soil or rock fills, 

and counterweight 

berms. 

Counter berm 

Figure 6.11 Butresses or Counter Berm for Slope 

Stabilsation 

6.7.4 

Soil Nailing 

Soil nailing is a method of in-situ reinforcement utilizing passive inclusions thatt will be mobilized if 

movement occurs. It can be used to retain excavations and stabilize slopes by creating 

in-situ, 

reinforced, soil retaining structures. 

6-14 

March 2009


Steel plate 

Soil face 

Shotcrete facing 

Main reinforcement 

Figure 6.12 Typical Details of Soil Nail 

6.7.5 

Geo-Synthetically Reinforcements 

Geo-synthetic soil reinforcement, such as geo-grid and geotextile, is another 

technique used to 

stabilize 

slopes. For high embankment on 

soft ground, the application of geo-synthetic i. .e., high 

strength geotextile or geogrid is 

required at 

the base of 

the embankment to enhance the stability of 

the embankment. 

6.7.6 

Retaining Walls 

The most common use of retaining walls for slope stabilization is when cut or 

fill is required and 

there is 

not sufficient space or right-of-way available for just the slope itself. Gravity and cantilever 

retaining walls are most common adopted. Examples of wall used are reinforced concrete wall, 

sheet pile wall, gabions wall, crib walls. 

March 2009 

6-15


REFERENCES 




New York, 1992, 1004 p. 





London 

[6] CGS, “Canadian Foundation Engineering Manual”, (Third edition). Canadian Geotechnical 









[11] Duncan, J. M., Buchignani, A. L., and DWet, M., An Engineering Manual for Slope Stability 

Studies, Department of Civil Engineering, Geotechnical Engineering, Virginia Polytechnic Institute 

and State University, Blacksburg, VA, 1987. 



[13] EM 1110-2-1902. Engineering and Design of Slope Stability, U.S. Army Corp of Engineer, 

[14] GCO (1984). Geotechnical Manual for Slope”. (Second Edition). Geotechnical Control Office, 

Hong Kong 

[15] GCO (1990) Review of Design Method for Excavation, Geotechnical Control Office, Hong 

Kong 


New Jersey 

[17] Huang Y.H., Stability Analysis of Earth Slopes, Van Nostrand Reinhold, 1983. 



- 494. 

6-16 March 2009


[19] Lambe T.W. and Whitman R.V., "Soil Mechanics", John Wiley 8: Sons, 1969 

[20] McCarthy D.J., "Essentials of Soil Mechanics and Foundations". 

[21] Mesri G., discussion of "New Design Procedure for stability of Soft Clays". by Charles C. Ladd 


Paper 10664. April 1975. pp. 409 - 412. 



[23] Nakashima, E., Tabara, K. & Maeda, Y.C. (1985). Theory and design of foundations on 

slopes. Proceedings of Japan Society of Civil Engineers, no. 355, pp 41-52. (In Japanese). 





1974. 






pp 2.13-2.72. 



[29] Skempton A.W. and D.H. McDonald, "The Allowable Settlement of Buildings", Proc. Inst. Civil 

Eng., Vo1.5 Pt.3. 1956, pp. 727-784. 







[34] Huang Y.H., Stability Analysis of Earth Slopes, Van Nostrand Reinhold, 1983. 

March 2009 6-17



6-18 March 2009


APPENDIX 6.A 

WORKED EXAMPLE: SLOPE STABILITY 

A.1 Problem 

The worked example presented herein illustrates the application of stability analysis by way of 

the Fellenius method of slices to determine the factor of safety in terms of effective stresses. 

The related slope configuration is shown in Figure 6.13 below. 

Figure 6.13 Related Slope Configuration 

The applicable soil properties and strength parameters are given as follows: 

i. 

ii. 

iii. 

iv. 

Soil unit weight (above & below water table), γs 

Effective cohesion, c’ 

Effective angle of shearing resistance, φ’ 

The soil mass is divided into slices of 1.5m wide sing the 

expression below (Eqn. 3.1), the resulting factor of 

safety is established as follows. 

= 20 kN/m 3 

= 10 kN/m 2 

= 29° 

F = c'L a+tan' ∑Wcosα-ul 

∑ Wsinα 

(6.3) 

Solution: 

i. The weight of each slice, W = γsbh 

= 20 x 1.5 x h 

= 30h kN/m 

March 2009 6A-1


ii. 

The height of each slice is set off below the centre of the base, and the 

normal and tangential components, h cos α and h sin α respectively are 

determined graphically as shown in Figure 3.3. Thus: 

W cos α 

W sin α 

= 

= 

30h cos α 

30h sin α 

iii. 

The pore water pressure at the centre of the base of each slice is taken to be 

γwzw, where zw is the vertical distance of the centre point below the water 

table (Fig 3.3 refers). [Note: This procedure slightly overestimates the pore 

water pressure, which strictly should be γwze, where ze is the vertical 

distance below the point of intersection of the water table and the 

equipotential through the centre of the slice base. The error involved is 

however, on the safe side]. 

iv. From Figure 6.13, the overall arc length, La is calculated as 14.35m. v. 

The results are summarised in Table 6.2 below. 

Table 6.5 Summary of Results 

Slice No. 

1 

2 

3 

4 

5 

6 

7 

8 

h cos α 

(m) 

0.75 

1.80 

2.70 

3.25 

3.45 

3.10 

1.90 

0.55 

h sin α 

(m) 

- 0.15 

- 0.10 

0.40 

1.00 

1.75 

2.35 

2.25 

0.95 

u 

(kN/m 2 ) 

5.9 

11.8 

11.2 

18.1 

17.1 

11.3 

0 

0 

l 

(m) 

1.55 

1.50 

1.55 

1.10 

1.70 

1.95 

2.35 

2.15 

u.l 

(kN/m) 

9.1 

17.7 

25.1 

29.0 

29.1 

22.0 

0 

0 

17.50 8.45 14.35 132.0 

vi. 

Hence: 

Σ W cos α = 30 x 17.50 

Σ W sin α = 30 x 8.45 

Σ (W cos α - ul) = 525 – 132 

= 525 kN/m 

= 254 kN/m 

= 393 kN/m 

F = c' L a +tan' ∑Wcosα-ul 

∑ Wsinα 

= 10x14.35+(0.554x393) 

254 

= 143.5+218 

254 

= 1.42 

6A-2 March 2009


A.2 PROBLEM 

Figure 6.14 shows a slope stability analysis of an embankment on soft clay using a commercial 

software; SLOPE/W. The soil stratums are as illustrated in Figure 6.14. In order to increase the 

factor of safety, two layers of high strength geotextiles were adopted. For embankment on soft 

soils, undrained condition is adopted. 

Figure 6.14 Stability Analysis of an Embankment Uses SLOPE/W Software 

March 2009 6A-3



6A-4 March 2009

CHAPTER 7 RETAINING WALL

Chapter 7 RETAINING WALL 


Table of Contents…………………………………………………………………………………………...…7-i 

List of Tables ........................................................................................................................ 7-II 

List of Figures ....................................................................................................................... 7-II 

7.1 GENERAL ..................................................................................................................... 7-1 

7.2 TYPE OF RETAINING WALLS .......................................................................................... 7-1 

7.3 SHEAR STRENGTH – LATERAL EARTH PRESSURE RELATIONSHIP ..................................... 7-2 

7.4 LATERAL EARTH PRESSURE ........................................................................................... 7-4 

7.4.1 At-Rest Lateral Earth Pressure .......................................................................... 7-4 

7.4.2 Active and Passive Lateral Earth Pressures ........................................................ 7-5 

7.4.2.1 Rankine’s Theory ............................................................................ 7-5 

7.4.2.2 Coulomb’s Theory ........................................................................... 7-8 

7.4.2.3 Effects of Wall Friction ..................................................................... 7-9 

7.4.3 Lateral Earth Pressure Due to Ground Water ................................................... 7-14 

7.4.4 Lateral Pressure from Surchage ...................................................................... 7-14 

7.5 STABILITY OF RIGID RETAINING WALL ....................................................................... 7-17 

7.5.1 Sliding/Translational Stability ......................................................................... 7-19 

7.5.2 Overturning Stability...................................................................................... 7-19 

7.5.3 Bearing Capacity Failure ................................................................................ 7-20 

7.5.4 Global Stability .............................................................................................. 7-20 

7.5.5 Selection of Backfill Materials ......................................................................... 7-21 

7.5.6 Design Wall Drainage System ......................................................................... 7-21 

7.5.7 Design Example – Gravity/Cantilever Reinforced Concrete Wall ......................... 7-23 

7.6 FLEXIBLE WALL SYSTEM ............................................................................................. 7-25 

7.6.1 General ........................................................................................................ 7-25 

7.6.2 Types of Flexible Walls .................................................................................. 7-26 

7.6.3 Sheet Pile Wall .............................................................................................. 7-27 

7.6.3.3 Design of Anchor - General ............................................................ 7-30 

7.6.3.4 Some Considerations on Sheet Pile Wall Design ............................... 7-31 

7.6.3.3 Cantilever Steel Sheet Pile Retaining Wall - Example ....................... 7-33 

REFERENCES........................................................................................................................ 7-38 

March 2009 7-i




7.1 Wall Displacements Required to Develop Active and Passive Earth Pressures 

(Wu, 1975) 7-5 

7.3 Calculation Table 7-24 

7.4 Permissible Steel Stress of Sheet Pile 7-32 



7.1 Forces Acting On Retaining Wall And Common Terminology 7-1 

7.2 Type of Retaining Walls 7-2 

7.3 State of Stress on a Soil Element Subjected to Stresses Induced by Wall 

Deformation 7-3 

7.4 The relationship between Ka, Kp, and Ko 7-4 

7.5 Development of Rankine Active and Passive Failure Zones for a Smooth Retaining 

Wall 7-6 

7.7 Schematic Of Coulomb’s Theory Plane Failure Wedge of Soil 7-8 

7.8 Comparison of Plane and Log-Spiral Failure Surfaces 7-10 

7.9 Passive Coefficients for Sloping Wall with Wall Friction and Horizontal Backfill 7-11 

7.10 Passive Coefficients for Vertical Wall with Wall Friction and Sloping Backfill 7-12 

7.11 Lateral Pressure Coefficient Chart for Granular Soil with Sloping Backfill 7-13 

7.12 General Distribution of Combined Active Earth Pressure and Water Pressure 7-14 

7.13 Lateral Pressure Due to Surcharge Loadings (after USS Steel, 1975) 7-16 

7.14 Potential Failure of a Rigid Retaining Wall 7-17 

7.15 Design Criteria for Rigid Retaining Walls (NAVFAC 1986) 7-18 

7.16 Typical Mode of Global Stability 7-20 

7.17 Potential Source of Subsurface Water 7-22 

7.19 Determining the Maximum and Minimum Pressures under the Base of the 

Cantilever Retaining Wall 7-23 

7.20 Typical Failure Mode of a Flexible Wall 7-25 

7.21 Type of Sheet Pile Walls 7-27 

7.22 Lateral Pressures Distribution for Fixed-End Method of Design of Cantilever 

Sheet Pile Wall in Granular Soils 7-29 

7.24 Various types of Anchoring for sheet pile walls 7-31 

7-ii March 2009


7 RETAINING WALL 

7.1 GENERAL 

Generally the main application of retaining wall is to hold back earth and maintain a difference in 

the elevation of the ground surface. The retaining wall is designed to withstand the forces exerted 

by the retained ground or “backfill” and other externally applied loads without excessive 

deformation or movement, and to transmit these forces safely to a foundation and to a portion of 

the restraining elements, if any, located beyond the failure surface. Figure 7.1 illustrated the forces 

acting on a retaining wall and some of the related terminology commonly used in retaining wall 

design. 

Special considerations are often necessary for retaining walls to be constructed close to land 

boundaries, particularly in urban areas. Land take requirement for construction often place 

limitations on the use of certain forms of earth retention. The cost of constructing a retaining wall is 

usually high compared with the cost of forming a new slope. Therefore, the need for a retaining 

wall should be assessed carefully during design. 

Figure 7.1 Forces Acting On Retaining Wall And Common Terminology 

7.2 TYPE OF RETAINING WALLS 

The rigidity or flexibility of a wall system is fundamental to the understanding of the development of 

earth pressures and the analysis of the wall stability. In simple terms, a wall is considered to be rigid 

if it moves as a unit in rigid body rotation and/or translation and does not experience bending 

deformation. Most gravity walls can be considered rigid walls. Flexible walls are those that undergo 

bending deformations in addition to rigid body motion. Such deformations result in a redistribution of 

lateral pressures from the more flexible to the stiffer portions of the system. Virtually all wall 

systems, except gravity walls, may be considered to be flexible. 

March 2009 7-1


Some of the typical retaining walls are as shown in Figure 7.2 

Cantilever 

Gravity Element 

Braced 

Tied-back (Anchored) 

Sheet Piling 

Counterfort wall 

Sheet Pile Wall 

Reinforced Soil 

Soil Nailing 

Figure 7.2 Type of Retaining Walls 

7.3 SHEAR STRENGTH – LATERAL EARTH PRESSURE RELATIONSHIP 

The concept of lateral pressure is related to the effective stress and shear strength discussed in 

Chapter 3, Item 3.2 to 3.4. It is recommended that reader should review the principles of effective 

stress shear strength before proceeding further in this Chapter. 

7-2 March 2009


The concept of lateral earth pressure acting on a wall can be explained based on the basic of the 

wall deformation. Consider an element of soil within a dry coarse-grained cohesionless soil mass. 

The geostatic effective stress on an element at any depth, z. would be as shown in Figure 7.3(a). 

Since the ground is not disturbed without any deformation, it is regarded as ‘at-rest’ condition. The 

coefficient of lateral pressure for this condition is termed as K0. 

Assume that a hypothetical, infinitely thin, infinitely rigid “wall” is inserted into the soil without 

changing the “at rest” stress condition in the soil as shown in Figure 7.3 (b). Now suppose that the 

hypothetical vertical wall move slightly to the left, i.e., away from the soil element as shown in 

Figure 7.3(c). In this condition, the vertical stress would remain unchanged. However, since the 

soil is cohesionless and cannot stand vertically on its own, it actively follows the wall. In this event, 

the horizontal stress decreases, which implies that the lateral earth pressure coefficient is less than 

Ko since the vertical stress remains unchanged. When this occurs the soil is said to be in the 

“active” state. The lateral earth pressure coefficient at this condition is called the “coefficient of 

active earth pressure”, Ka. 

δ a 

δ p 

p o p o p o p o 

p h =K o p o 

p h =K o p o p h =K a p o p h =K p p o 

Figure 7.3 State of Stress on a Soil Element Subjected to Stresses Induced by Wall Deformation (a) 

In-situ vertical and horizontal stresses (b) Insertion of hypothetical infinitely thin and infinitely rigid 

(c) Active contition of wall movement away from retained soil (d) Passive contition of wall 

movement toward retained soil 

Now, instead of moving away from the soil, suppose the hypothetical vertical wall move to the right 

into the soil element as shown in Figure 7.3 (d). Again, the vertical stress would remain unchanged. 

However, the soil behind the wall passively resists the tendency for it to move, i.e., the horizontal 

stress would increase, which implies that the lateral earth pressure coefficient would become 

greater than Ko since the vertical stress remains unchanged. When this occurs the soil is said to be 

in the “passive” state. The lateral earth pressure coefficient at this condition is called the “coefficient 

of passive earth pressure,” Kp. 

The relationship between Ka, Kp, and Ko can best be illustrated graphically by Figure 7.4 below. 

March 2009 7-3


K 

Kp K p ((Passive limit) 

limit) 

Ko K o ((at at rest) 

) 

( (Not not a failure limit) 

) 

Ka K a ((Active active limit) 

limit ) 

δ 

δ, , lateral lateral soil soil movement 

movement 

7.4 LATERAL EARTH PRESSURE 

7.4.1 At-Rest Lateral Earth Pressure 

Figure 7.4 The relationship between Ka, Kp, and Ko 

The at-rest earth pressure condition in Figure 7.3(a) and (b) represents the lateral effective stress 

that exists in a natural soil in its undisturbed state. For cut walls constructed in near normally 

consolidated soils, the at-rest earth pressure coefficient, Ko, can be approximated by the equation 

(Jaky, 1944): 

K o = 1 – sin φ′ (7.1) 

where φ′ is the effective (drained) friction angle of the soil. 

The magnitude of the at-rest earth pressure coefficient is primarily a function of soil shear strength 

and degree of overconsolidation, which, as indicated in Chapter 4, may result from natural geologic 

processes for retained natural ground or from compaction effects for backfill soils. In 

overconsolidated soils, K o can be estimated as (Schmidt, 1966): 

K o 

= (1 − sin ′)(OCR) Ω (7.2) 

where Ω is a dimensionless coefficient, which, for most soils, can be taken as sin φ′ (Mayne and 

Kulhawy, 1982) and OCR is the overconsolidation ratio. Typical values of K0 are as shown below: 

Normally consolidated clay, Ko = 0.55 to 0.65 

Lightly overconsolidated clays (OCR ≤ 4) Ko = up to 1 

Heavily overconsolidated clays (OCR > 4) Ko = > 2 (Brooker and Ireland, 1965) 

Sand Ko = 0.4 to 0.5 

7-4 March 2009


At-Rest condition may be appropriate for heavily preloaded, stiff wall systems. However, at-rest 

conditions are not typically used for flexible wall systems such as steel sheet-pile wall, where the wall 

undergoes some lateral deformation and designing to a requirement of zero movement is not 

practical. 

7.4.2 Active and Passive Lateral Earth Pressures 

Active earth pressure (condition in Figure 7.3(c)) occurs when the wall moves away from the soil 

and the soil mass stretches horizontally sufficient to mobilize its shear strength fully, and a condition 

of plastic equilibrium is reached. The ratio of the horizontal component or active pressure to the 

vertical stress is the active pressure coefficient Ka. 

Passive earth pressure occurs when a soil mass is compressed horizontally, mobilizing its shear 

resistance fully. The ratio of the horizontal component of passive pressure to the vertical stress is 

the passive pressure coefficient, Kp. 

The amount of movement necessary to reach the plastic equilibrium conditions is dependent 

primarily on the type of backfill material. Some guidance on these movements is given in Table 7.1 

Table 7.1 Wall Displacements Required to Develop Active and Passive Earth Pressures 

Soil Type and Condition 

Necessary Displacement 

Active 

Passive 

Dense Cohesiveless 0.001H 0.02H 

Loose Cohesiveless 0.004H 0.06H 

Siff Cohesive 0.01H 0.02 H 

Soft Cohesive 0.02H 0.04H 

Note : H = Wall Height 

(Source: Wu, 1975) 

There are two well-known classical lateral earth pressure theories i.e. Rankine’s and Coulomb’s. 

Each furnishes expressions for active and passive pressures for a soil mass at the state of failure. 

7.4.2.1 Rankine’s Theory 

Rankine’s Theory is based on the assumptions that the wall introduces no changes in the shearing 

stresses at the surface of contact between the wall and the soil. It is also assumed that the ground 

surfaces is a straight line (horizontal or inclined straight line) and that a plane failure surface 

develops. 

March 2009 7-5


Figure 7. 5 Development of Rankine 

Active and Passive Failure Zones for a 

Smooth Retaining Wall 

When the 

Rankine state of failure has been reached, active and passive failure zones will develop as 

shown in 

Figure 7.5. The coefficient of active 

and passive 

earth pressure are expressed by the 

following 

equations: 

- - 

- 

(7.3) 

- 

- - 

(7.4) 

Where 

= the sloping angle of the backfill behind the wall 

a = the active 

earth pressure coefficient 

p = the passive earth preesure coefficient 

= the effective frictional angle of the soil 

K a 

K p 

φ Note that for the case 

of cohesionless soil on level backfill, thesse equations are reduced to 

Ka = 

- 

tan 2 (45 - 

) 

(7.5) 

Kp = 

- 

tan 2 (45 - 

) 

(7.6) 

Thus, without considering the ground water level, the distribution of lateral earth pressures can be 

assumed 

to be triangnular (see Figure 7.6) such 

that 

7-6 

March 2009


p a = K a p 0= K a γ z (7.7) 

p a = K p p 0 = K p γ ζ (7.8) 

where 

p 0 

= Effective overburden pressure (unit length)= γh 

pa = Active lateral earth pressures (unit length) 

pp = Passive lateral earth pressures (unit length) 

z = Depth below the ground surface 

h = Depth of tension crack (clayey soil only) 

Z 

H 

p a =rZ tan 2 (45- Ø ) 2 

P p =rZ tan 2 (45+ Ø ) 2 

(a) 

Z 

ß 

ß 

p a =rZK o 

p p =rZK p 

K a = cosß β β 

β β 

K a = cosß β β 

β β 

K a = 1 K p 

Z 

2c tan (45°- Ø ) 

2c tan (45+ Ø ) 2 

2 

Z 

p a = rZ tan 2 (45- Ø )-2c 2 tan(45°-Ø) P p = rZ tan 2 (45+ Ø )+2c 2 tan(45+Ø) 

2 

2 

(b) 

Figure 7.6 Triangular Lateral Force Distribution By Rankine Theory (a) For Granular Soil (b) For 

Cohesive Soil With Tension Crack Depth ‘H’ (Active Case) 

For non- granular (c’ – φ ‘) soils, the lateral pressures are : 

P a = K a γz – 2cK a (7.9) 

P p = K p γz + 2cK p (7.10) 

c = Cohesive strength of soil 

Theoretically, in soils with cohesion, the active earth pressure behind the wall becomes negative 

from the ground surface to a critical depth z where γh is less than 2c′ √ K a . This critical depth is 

referred to as the “tension crack.” The active earth pressure acting against the wall within the depth 

of the tension crack is assumed to be zero. Unless positive drainage measures are provided, water 

infiltration into the tension crack may result in hydrostatic pressure on the retaining structure and 

should be full added to the lateral earth pressure. 

March 2009 7-7


7.4.2.2 Coulomb’s Theory 

Coulomb Theory is also based on limit equilibrium of a plane wedge of soil. However, the theory 

takes into consideration the effects of wall friction, sloping wall face as well as the sloping backfill. 

The pressures calculated by using these coefficients are commonly known as the Coulomb earth 

pressures. Since Coulomb’s method is based on limit equilibrium of a wedge of soil, only the 

magnitude and direction of the earth pressure is found. Pressure distributions and the location of the 

resultant are assumed to be triangular. Coulomb’s coefficients of lateral pressures are as follows with 

their related terms and pressures diagrams shown in Figure 7.7 

K a = 

cos 2 - θ 

(7.11) 

cos 2 sin- θ sin- β 

θ cosθ+ δ 

cos- δ cos- β 

cos 2 + θ 

K p = 

(7.12) 

cos 2 

θ cosθ - δ 

 

Figure 7.7 Schematic Of Coulomb’s Theory Plane Failure Wedge of Soil 

(a) Active Condition (b) Passive Condition 

7-8 March 2009


7.4.2.3 Effects of Wall Friction 

The magnitude and direction of the developed wall friction depends on the relative movement 

between the wall and the soil. In the active case, the maximum value of wall friction develops only 

when the soil wedge moves significantly downwards relative to the rear face of the wall. In some 

cases, wall friction cannot develop. These include cases where the wall moves down with the soil, 

such as a gravity wall on a yielding foundation or a sheet pile wall with inclined anchors, and cases 

where the failure surface forms away from the wall, such as in cantilever and counterfort walls. 

The maximum values of wall friction may be takes as follows : 

Timber, steel, precast concrete wall 

Cast in-situ concrete wall 

δ max. = Ø’/2 

δ max. = 2 Ø’/3 

Considerable structural movements may be necessary, however, to mobilize maximum wall friction, 

for which the soil in the passive zone needs to move upwards relative to the structure. Generally, 

maximum wall friction is only mobilized where the wall tends to move downwards, for example, if a 

wall is founded on compressible soil, or for sheet piled walls with inclined tensioned members. 

Some guidance on the proportion of maximum wall friction which may develop in various cases is 

given below (Teng) 

δ = 20 0 concrete or brick walls 

= 15 0 uncoated sheetpile 

= 0 0 if wall tends to move downward together with the soil 

= 0 0 sheetpiling with small penertration or penetrated into soft or loose soil 

= 0 0 if backfill is subjected to vibratiion 

In general, the effects of wall friction on Rankine and Coulomb methods of earth pressure 

computation are as follows: 

a) The Rankine method cannot take account of wall friction. Accordingly, K a is overestimated 

slightly and K p is under-estimated, thereby making the Rankine method conservative for 

most applications. 

b) The Coulomb theory can take account of wall friction, but the results are unreliable for 

passive earth pressures for wall friction angle values greater than φ′/3 because the failure 

surface is assumed to be a plane. The failure wedges assumed in the Coulomb analysis take 

the form of straight lines as shown in Figure 7.8. However, this contrasted with the curved 

shapes of failure surface observed in many model tests. This assumption resulted in K a 

being underestimated slightly and K p being overestimated very significantly for large values 

of φ′. 

In general, the effect of wall friction is to reduce active pressure. It is small and often disregarded. 

However, wall friction increases the value of K p significantly and thus could yield lateral earth 

pressure that could be very large and could be unsafe as passive earth pressure forces are generally 

resisting forces in stability analysis 

March 2009 7-9


Figure 7.8 Comparison of Plane and Log-Spiral Failure Surfaces (a) Active Case (b) Passive Case 

Hence, it is recommended that the log-spirall failure surface (shown 

resemblee more closely the actual failure plane be used to calculate 

coefficients. 

in Figure 7.8) which could 

the passive earth pressure 

Charts for two common wall configurations, sloping wall with level backfill and vertical wall with 

sloping backfill based 

on the log-spiral theory are presented in Figures 

7.9 and 7.10 (Caquot and 

Kerisel, 1948; NAVFAC, 1986b). For walls that have a sloping backface and sloping backfill, the 

passive earth pressuree coefficient can be calculated as indicated in Figure 

7.9 and 7.10 by using δ = 

′/3. For granular soils, the coefficients of earth pressure can be deived from Figure 7.11 

7-10 

March 2009


Figure 7.9 Passive Coefficients for Sloping Wall with Wall Friction and Horizontal Backfill 

(Caquot and Kerisel, 1948; NAVFAC, 1986b) 

March 2009 

7-11


Figure 7.10 Passive Coefficients for Vertical Wall with Wall Friction and Sloping Backfill 

(Caquot and Kerisel, 1948; NAVFAC, 1986b) 

7-12 

March 2009


Figure 7.11 Lateral Pressure Coefficient Chart for Granular Soil with Sloping Backfill 

March 2009 7-13


7.4.3 

Lateral Earth Pressure Due to Ground Water 

In cases where ground water exists, the lateral pressure due to the water at any depth below the 

ground water level is equal to the hydrostatic pressure at that point since the friction angle of water 

is zero and use of either Equation 

7.5 or 7.6 leads to a coefficient of lateral pressure for water, Kw 

equal to 

1.0. The computation of the vertical water pressure is based on triangular pressure 

distribution that increases linearly with depth as illustrated in Figure 7.12. The lateral earth pressure 

is added to the hydrostatic water pressure to obtain the total lateral pressure acting on the wall at 

any point below the ground water level. For a typical soil 

friction angle of 30 degrees, Ka = 1/ /3. 

Since Kw = 1, it can be seen that the lateral pressure due to 

water is approximately 3 times that due 

the active lateral earth pressure. Thus, it is important to provide adequate drainage behind the wall 

to reducee and control the ground water table build-up. 

Figure 7.12 General Distribution of Combined Active Earth Pressuree and Water Pressure 

7.4.4 

Lateral Pressure from Surchage 

Surcharge loads on the backfill surface near an 

earth retaining structure also cause lateral pressures 

on the structure. The 

loading cases usually consist of: 

• Uniform surcharge 

• Point 

loads 

• Line loads parallel to the wall 

• Strip 

loads parallel to the wall. 

Surcharge loads (vertical loads applied at the ground surface) are assumed to result in a uniform 

increase in lateral pressure over the entire height of the 

wall. The uniform increase in lateral 

pressure for a uniform 

surcharge loading can be 

written as: 

7-14 

March 2009


∆p s 

= K q s 

(7.13) 

where ∆ps 

qs 

K 

= increase in lateral earth pressure due to the vertical surcharge load 

= vertical surcharge load applied at the ground surface, 

= appropriate earth pressure coefficient. 

When traffic is expected to come to within a distance from the wall face equivalent to one-half the 

wall height, the wall should be designed for a live load surcharge. The standard loadings for 

highway structures in are expressed in terms of HA and HB loading as defined in BS 5400 : Part 2 : 

1978. In the absence of more exact calculations, the nominal load due to live load surcharge may 

be taken from Table 7.2. 

Table 7.2 Suggested Surcharge Loads to be Used in the Design of Retaining Structures 

Road class 

Type of live loading 

Equivalent 

surcharge 

Urban trunk 

Rural trunk 

(Road likely to be regularly used by 

HA + 45 units of HB 

20kPa 

heavy industrial traffic) 

Primary distributor 

Rural main road 

HA = 37 ½ units of HB 

15kPa 

District and local distributors 

Other rural roads 

HA 

10kPa 

Access Roads, Carparks 

Footpaths, isolated from roads 

5kPa 

Play areas 

Note : 1. It is recommended that these surcharges be applied to the 1 in 10 year storm condition. 

2. For footpaths not isolated from roadways, the surcharge applying for that road class 

should be used. 

(Source: Public Works Department, 1977) 

Point loads, line loads, and strip loads are vertical surface loadings that are applied over limited areas 

as compared to surcharge loads. Hence, the increase in lateral earth pressure used for wall system 

design is not constant with depth as is the case for uniform surcharge loadings. These loadings are 

typically calculated by using equations based on elasticity theory for lateral stress distribution with 

depth and are as shown in Table 7.13. Lateral pressures resulting from these surcharges should be 

added explicitly to other lateral pressures. 

March 2009 7-15


Figure 7.13 Lateral Pressure Due to Surcharge Loadings (after USS Steel, 1975) 

7-16 March 2009


7.5 

STABILITY OF RIGID RETAINING WALL 

Rigid retaining walls are those that develop their lateral resistance primarily from their own weight 

and the weight of soil above the base of the wall, if any. The 

goetechnical design analysis for a rigid 

retaining 

wall shall include all the possible mode 

of a rigid retaining wall, namely 

a) Sliding/translational failure 

b) Rotational failure 

c) Foundation bearing capacity failure 

d) Deep seated/global stability failure 

Figure 7. .14 shows the 

schematic sketch of the potential failures of a rigid retaining wall. 

(a) Sliding or translational failure 

(b) Rotational failure 

(c) Bearing Capacity failure 

d) Deep-seated Failure 

Figure 7.14 

Potential Failure of a Rigid Retaining Wall 

The stability of free standing rigid retaining wall can be determined by computing factors of safety, 

which may be deined in general equation as: 

The forces that produce overturning and sliding 

also produce 

the foundation bearing pressures and, 

therefore, (a), (b) and 

(c) are interlated for most soils 

Figure 7.15 presented a useful guide for the 

computation of the stability of a rigid concrete 

retaining 

wall (after NAVFAC, 1986). 

March 2009 

7-17


Definitions 

B = width of the base of the footing 

tan δ t = friction factor between soil and base 

W = weight at the baseof wall. Includes 

weight of wall for gravity walls. Includes 

weight of the soil above footing for 

cantilever and counterfort walls 

c = cohesion of the foundation soil 

c a = adhesion between concrete and soil 

δ = angle of wall friction 

= passive resistance 

P p 

Location of Resultant, R 

Based on moments about toe (assuming P p =0) 

d = Wa+P vg-P h b 

W+P v 

Criteria for Eccentricity, e 

e = d- B ; e≤B/6 for soils; e≤B/4 for rocks 

2 

Factors of Safety Against Sliding 

FS δ = W+P v tan δ b +c a B 

≥1.5 min 

P h 

Applied Stress at Base (q max , q min , q eq ) 

q max = W+P v 

(1+ 6e 

B 

q min = W+P v 

B 

B ) 

(1- 6e 

B ) 

Equivalent uniform (Meyerhof) applied stress, q eq 

is given as follows: 

q eq = W+P v 

where B’ = B-2e 

B' 

Use uniform stress, q eq , for soils and settlement 

analysis; use trapezoidal distribution with q max 

and q min for rocks and structural analysis 

Deep-seated (Global) Stability 

Evaluate global stability using guidance in Chap. 

6 (Slope Stability) 

Figure 7.15 Design Criteria for Rigid Retaining Walls (NAVFAC 1986) 

7-18 March 2009


7.5.1 Sliding/Translational Stability 

The horizontal component of all lateral pressures tends to cause the wall to slide along the base of 

the wall (or along any horizontal section of a gravity and crib wall). If the passive resistance is 

neglected, the sliding force along the bottom of the wall is resisted by a horizontal force which 

consists of friction, adhesion or a combination of both. If the bottom of base slab is rough, as the 

case of concrete poured directly on soil, the coefficient of friction is equal to tan φ', (φ' is the angle 

of internal friction of the soil). Typical coefficients of friction are as follows: 

Course-grained (without silt) 0.55 

Course-grained (with silt) 0.45 

Silt 0.35 

Sound rock (with rough surface) 0.60 

For cohesive soils the adhesion between the base slab and the soil is assumed to be equal to the 

cohesive strength of the clay and φ is assumed to be zero. The designer should consider the 

possibility of reduction in cohesive strength due to construction works such as excavation, exposure 

to surface water etc. If the retaining wall is supported on piles, the entire vertical and horizontal load 

should be assumed to be carried by piles. No frictional resistance and no adhesion should be 

assigned along the base slab. 

For checking the sliding factor of safety, the live load surcharge is usually not considered in the 

stabilising forces over the heel of the wall. Also, the passive resistance of the soil in front of the wall 

is commonly neglected in the stability analysis. If it is included in the computation, as in the case 

where the toe of wall is covered by a large depth of soil, its value should be reduced to take care of 

the high potential of the soil to be removed by erosion, future excavation, and tension cracks in 

cohesive soils. 

The minimum safety factor for sliding/translational stability shall be of minimum 1.5. The sliding 

stability can be increase by either increasing the overall weight of the retaining wall or providing 

sufficient passive lateral resistance of the wall. This can be done by introducing a wider base, 

construction of structural shear key and incorporating deep foundation support. 

7.5.2 Overturning Stability 

The lateral pressure due to the backfill and surcharge tends to tip the retaining over about its toe. 

This overturning moment is stabilised by the weight of the wall and the weight of the soil above the 

base of the wall. The overturning stability of the wall is always the most critical potential mode of 

failure when the walls are underlain by weak soils. The minimum factor of safety against overturning 

is: 

F s = 

Sum of stabilizing moment 

Sum of overturning moment 

≥2.0 

To overcome the overturning stability, normally pile foundation is recommended. For some cases, 

ground improvement such as removal and replacement is adopted to increase the bearing capacity of 

the ground (provided the soft bearing ground is relatively thin). 

For passive resistance of the soil in front of the wall, designer should evaluate whether to ignore or 

to use a reduced value basing on the reason discussed in 7.5.1 above. 

March 2009 7-19


7.5.3 Bearing Capacity Failure 

The computed vertical pressure at the base of the wall footing must be checked against the ultimate 

bearing capacity of the soil. The generalized distribution of the bearing pressure at the wall base is 

illustrated in Figure 7.15. Note that the bearing pressure at the toe is greater than that at the heel. 

The magnitude and distribution of these pressures are computed by using the applied loads shown in 

Figure 7.15. The equivalent uniform bearing pressure, q eq , should be used for evaluating the factor 

of safety against bearing capacity failure. The procedures for determining the allowable bearing 

capacity of the foundation soils can be found in Chapter 5 (Bearing Capacity) of this Volume. 

Generally, the factor of safety against bearing failure is defined as 

Where 

F s = q ult 

q eq 

≥ 2.0 

q ult = ultimate bearing pressure 

q eq = equivalent uniform bearing pressure (as computed according to Figure 10.15) 

7.5.4 Global Stability 

The overall stability shall be checked to avoid deep seated failure due to circular rotational or noncircular 

failure beyond the retaining wall. It must be checked with respect to the most critical failure 

surface. The minimum factor of safety for the overall stability shall be of minimum 1.5. A typical 

mode of circular rotational stability condition is illustrated in Figure 7.16 

If global stability is found to be a problem, deep foundations or the use of lightweight backfill may be 

considered. Alternatively, measures can be taken to improve the shear strength of the weak soil 

stratum. Other wall types, such as an anchored soldier pile and lagging wall or tangent or secant 

pile wall, should also be considered in this case. 

Figure 7.16 Typical Mode of Global Stability 

7-20 March 2009


7.5.5 Selection of Backfill Materials 

The ideal backfill for a retaining is a free draining granular material of high shearing strength. 

However, the final choice of material should be based on the costs and availability of such materials 

balanced against the cost of more expensive walls. 

In general, the use of fine-grained clayey backfills is not recommended due to the following 

reasons: 

a) Clays are subject to seasonal variations in moisture content and consequent swelling and 

shrinkage. This effect may lead to an increase in pressure against a wall when these soils 

are used as backfill. 

b) As clays are subjected to consolidation, long terms settlement problems are considerably 

greater than with cohesionless materials. 

c) For clay backfill, special attention must be paid to the provision of drainage to prevent the 

build-up of water pressure. Free draining cohesionless materials may not require the same 

amount of attention in this respect. 

d) The wall deflection required to produce the active state in cohesive materials with a 

significant clay content may be up to 10 times greater than for cohesionless materials. 

This, together with the fact that the former generally have lower values of shearing 

strength, means that the amount of shear strength mobilized for any given wall movement 

is considerably lower for cohesive materials than for cohesionless materials. The 

corresponding earth pressure on the active side for a particular wall movement will 

therefore be higher if cohesive soil is used for backfill. 

It is essential to specify and supervise the placing of backfill to ensure that its strength and unit 

weight properties agree with the design assumptions both for lateral earth pressure and dead 

weight calculations. In this regard, it is particularly important to ensure that the backfill behind a 

wall and on a slope is properly compacted. The backfill should normally be compacted in thin layers 

using light compaction plant so as not to minimize compaction loading on the wall. 

7.5.6 Design Wall Drainage System 

Control of water is a key component of the design of earth retaining structures. Both subsurface 

water and surface water can cause damage during and/or after construction of the wall. Surface 

water runoff can destabilize a structure under construction by inundating the backfill. It can also 

destabilize a completed structure by erosion or by infiltrating into the backfill. Hence, adequate and 

proper design for surface water runoff is important to ensure the stability of the wall. Potential 

sources of subsurface water are surface water infiltration and groundwater as illustrated in Figure 

7.17. 

March 2009 7-21


Surface Water 

Infiltration 

Drainage 

aggregate 

Fill 

Retained Fill 

Groundwater 

Foundation Soil 

Figure 7.17 Potential Source of Subsurface Water 

Drainage system design depends on wall type, backfill and/or retained soil type, and groundwater 

conditions. Drainage system components such as granular soils, prefabricated drainage elements and 

filters, are usually sized and selected based on local experience, site geometry, and estimated flows, 

although detailed design is only occasionally performed. Drainage systems may be omitted if the wall 

is designed to resist full water pressure. 

Drainage measures for fill wall systems and cut wall systems typically consist of the use of a freedraining 

material at the back face of the wall, with “weep holes” and/or longitudinal collector drains 

along the back face as shown in Figure 7.18. The collector drains may be perforated pipes or gravel 

drains. Where weepholes are used, BS 8002 specified that they should be at least 75 mm in diameter 

and at a spacing of not more than 1 m horizontally and 1 m to 2 m vertically. 

7-22 March 2009


Wall Backfill 

Face chimney 

drain 

Retained Backfill 

Chimney 

drain 

Weephole 

Collection & 

Drain Pipes 

Outlet Pipe 

Figure 7.18 Some Typical Retaining 

Wall Drainage 

7.5.7 

Design Example – Gravity/Can 

ntilever Reinforced Concrete Wall 

Determine the maximum and minimum pressures under the 

base of the cantilever retaining wall as 

shown in 

Figure 7.19 below, and the factor of safety against sliding. 

Figure 7. .19 Example Calculation for Stability of a Cantilever Retaining Wall 

March 2009 

7-23


The applicable soil properties and strength parameters are given as follows: 

Soil unit weight, γ s = 17 kN/m 3 

Effective cohesion, c’ = 0 kN/m 2 

Effective angle of shearing resistance, φ’ = 40 o 

Assume friction on the base of wall, δ = 30 o 

Unit weight of concrete, γ c = 23.5 kN/m 3 

And, water table is below base of wall. 

Solution: 

i. To determine the position of the base reaction, the moment of all forces about the heel of 

the wall (X) are calculated as follows (Table 7.3 refers). 

Table 7.3 Calculation Table 

Force per m (kN) Arm (m) Moment per m 

(kNm) 

(1) 0.22 x 40 x 5.40 = 47.5 2.70 128.2 

(2) ½ x 0.22 x 17 x 5.40 2 = 54.6 

R h = 102.1 

1.80 98.3 

(Stem) 5.00 x 0.30 x 23.5 = 35.3 1.90 67.0 

(Base) 3.00 x 0.40 x 23.5 = 28.2 1.50 42.3 

(Soil) 5.00 x 1.75 x 17 = 148.8 0.875 130.2 

(Load) 1.75 x 40 = 70.0 

R v = 282.3 

0.875 61.3 

M = 527.3 

The active pressure is calculated on the vertical through the heel of the wall. No shear stresses act 

on this vertical, and therefore the Rankine theory (δ = 0) is used to calculate the active pressure 

using the pressure distribution as shown in Figure 1 above. Thus: 

For φ’ = 40 0 (and δ = 0), K a = 0.22 

Lever arm of base resultant, M R v 

= 

527.3 

282.3 

= 1.81 

i.e., the resultant acts within the middle third of the base. 

ii. Thus, eccentricity of base reaction, e = 1.81 – 1.50 

= 0.31 m 

The maximum and minimum base pressures are given by: 

R v 6e 

1± 

B B 

p = 282.3 

3 

1± 6x0.36 

= 94 (1 ± 0.72) 

B3 

= 112 kN/m 2 and 21 kN/m 2 

7-24 March 2009


Thus the 

factor of safety against sliding is given 

by: 

F = = 

= 1.1 ≤ 1.5 

not OK, need to increase resistance against sliding either 

by increasing 

the width of the base 

slab, introduce shear key 

or using raked pile. 

7.6 

7.6.1 

FLEXIBLE WALL SYSTEM 

General 

Unlike rigid retaining wall, the stability of the flexible wall depends mainly on the embedded length 

of the wall element. Some of the common types of flexible wall are sheet 

pile wall, soldier pile wall, 

contiguous bored pile wall and diaphragm wall. Sometimes due to stability requirement, tie backs or 

anchors to deadman and strut system are used to increase the overall stability of the wall. 

The common failure modes of a flexible retaining wall are: 

a) Rotational failure (at strut/ /tie back or at 

toe of the wall) 

b) Deep seated/global stability failure 

c) Hydraulic failure due to piping and uplift (in case of high differential hydrostatic head) 

d) Structural failure (tie back failure or wall element failure) 

(a) Deep-seated failure 

(b) Rotation about the anchor/prop 

(c) Rotation near base 

(d) Failure of 

(e) Failure by bending 

Figure 7.20 Typical Failure Mode of a Flexible Wall 

March 2009 

7-25


7.6.2 Types of Flexible Walls 

The following retaining wall types are commonly used in Malaysia either to retain and/or support 

soils during excavations: 

a) Sheet pile wall 

b) Soldier pile wall 

c) Contiguous bored pile / caisson wall 

d) Diaphragm wall 

a) Sheet Pile Walls 

The sheet pile wall is used in many types of temporary and permanent structures. It is one of the 

most common methods used in the Department especially for the support and protection of river 

banks, water front construction, flood defence as well as temporary supports or containment for 

construction of hydraulic structures. Steel sheet piles are preferred mainly because of their ease of 

installation, length of service life and ability to be driven through water. However, they are not 

suitable when high bedrock or boulders prevent penetration to the required depth. 

When selecting sheet piles to be used, it is important to consider the drivability of the piles. The 

ability of the sheet pile to penetrate the ground depends on the section size of the pile and the type 

of the pile hammer used, as well as the ground conditions. It is difficult to drive sheet piles through 

soils with Standard Penetration Test (SPT) ‘N’ values greater than 50 (subjected to pile section). 

Further discussion on the basic principles in design of sheet pile wall are discussed in Item 7,6.3 

below. 

b) Soldier Pile Wall 

Soldier pile wall has two basic components, soldier piles (vertical component) and lagging 

(horizontal component). Soldier piles provide intermittent vertical support and are installed before 

excavation commences. Due to their relative rigidity compared to the lagging, the piles provide the 

primary support to the retained soil as a result of the arching effect. Spacing of the piles is chosen 

to suit the arching ability of the soil and the proximity of any structures sensitive to settlement. A 

spacing of 2 – 3 m is commonly used in strong soils and no sensitive structures are present. The 

spacing is reduced to 1 – 2 m in weaker soils or near sensitive structures. 

c) Contiguous Bored Pile /Caisson Wall 

Replacement pile wall i.e., contiguous bored pile wall or caisson wall is the common excavation 

support system adopted in Malaysia. Generally, these types of wall are used as the permanent 

retaining wall system for basement construction and sometimes for high wall in hillside 

development. 

Bored piles or caisson piles are constructed continuously in a row to form retaining structures. A 

gap of approximately 75mm to 100mm is allowed between the piles. for ground with high ground 

water table or loose soils, grout columns are introduced between the gaps behind the wall system. 

For a better water tide conditions pressured grout columns can be used to minimize the water 

leakage. 

For caisson wall, it is commonly used at areas with limited working space; where big machinery i.e., 

boring rig and excavator are not possible. 

7-26 March 2009


d) Diaphragm Wall 

Diaphragm wall construction is very similar to bored pile wall. This wall system comes in panels and 

the soil removal is using a mechanical grab. Water stopping 

system is introduced between the wall 

panels to 

ensure total 

water tightness. 

Diaphragm wall system is not suitable for area with shallow bed rock. Rock chiseling during the 

installation may affect the construction duration and causing vibration disturbance to the 

surrounding. 

7.6.3 

7.6.3.1 

Sheet Pile Wall 

Types of Sheet Pile 

Wall 

The sheet pile wall system can be 

further divided into the 

followings categories according to the 

form of support provided, namely:- 

a) Cantilevered or unbraced wall 

b) Supported wall either with anchor/tie-back or bracing/struts 

The various types of sheet pile wall are as illustrated in Figure 7.21 

Figure 7.21 Type of Sheet Pile Walls 

a) Cantilever Sheet Pile Wall 

A cantilever sheet pile wall is one that does 

not have any additional support such as bracing, 

anchors, or other structural elements and thus relies on its flexural strength and embedment to resist 

the lateral earth pressures. The imposed lateral earth pressures on these walls create large flexural 

stresses in the steel and as such, 

these types of wall generally are not 

more than 3 to 4 m high. 

Cantilever walls also experience greater lateral deflections and are more susceptible to 

failure due to 

scour or erosion of the 

supporting soils. 

b) Supported Sheet Pile Wall 

Most sheet pile walls 

include additional lateral supports, using internally bracing/struts or tieback 

anchors (known as braced walls or anchored walls respectively). The additional support provided 

reduces the flexural stresses and lateral movements in the wall, thus permits construction of walls 

much taller than that of cantilever 

design. In this situation the soil conditions at the toe of the wall 

are not as critical to the overall stability of the structure and depth of embedment required would not 

be as deep as in the case of a cantilever wall. 

March 2009 

7-27


In general, a wall supported by a single tie or prop will generally will only be cost-effective up to 

a 

retained height of 10 

m. Also, as the wall does not move as much, there is less settlement in the 

backfill. When more than one level of supports are used, wall stability becomes a function of the 

support stiffness and the conventional active/passive earth pressure distribution does 

not necessary 

apply. 

7.6.3.2 

Design of Sheet Pile Wall 

In general, the design of sheet pile wall requires two sets of calculations, one to determine the 

geometry 

of the sheet 

pile to achieve equilibrium 

under the design conditions, the other to determine 

the structural requirements of the 

wall to resist the induced bending moments and shear forces 

derived from the equilibrium calculation. 

To design the steel sheet pile wall, several empirical and semi-empirical methods have been 

developed, all of which are based on the classical lateral earth pressuress theories. Several methods 

have been developed in the design 

of sheet pile wall; however the two most common 

methods are 

the Free-eninfluencee with which the depth of 

embedment 

has on the 

deflected shape of the wall. Only the 

basic concepts and lateral pressuree distribution 

are discussed below. Reader can refer to the many 

referencee books on the detailed design of sheet 

pile wall, among which are Piling Handbook, Arcelor 

Groups, ‘ ‘Foundation design’ by W.C. Teng and ‘Steel Sheet Piling Design Manual’, USS. 

a) Free-end method. 

The Free-end method 

is based on 

the assumption that the 

sheet pile is 

embedded to a sufficient 

depth into the soil to prevent translation, but not rotation at the toe 

and a pinned support is 

method and Fixed-end method. The main different between 

these methods lies in the 

assumed. This condition and the idealised earth pressure distribution are as shown in Figure 7.21. 

For the supported wall, a strut (prop) or tie near the top of the wall provides the other support. 

Compare 

to Fixed-endd method under similar set of conditions, the relative length of pile required is 

less but the maximumm moments are higher. 

(a) 

(b) 

Figure 7.21: Free-end Method of 

Design of Single Prop Sheet Pile Wall 

7-28 

March 2009


b) Fixed-end Method 

A wall designed using Fixed-end principles is embedded sufficiently deep enough so that at the foot 

of the wall, both translation and rotation are prevented and fixity is assumed. This is the condition 

assumed in the design of a cantilever sheet pile wall. Figure 7.22 (a) and (b) illustrated the 

deflected shape of a cantilever sheet pile together with the conventional and simplified pressure 

distributions used for design. An example on the application of this method in Cantilever sheet pile 

wall desiGn is given in Item 7.6.3.3 below. 

Dredge Line 

Deflected shape 

of pile 

(a) 

Figure 7.22 Lateral pressures distribution for Fixed-end Method of design of cantilever 

sheet pile wall in granular soils: (a) Idealized distribution (b) Simplified distribution 

(b) 

A tie or prop may also be provided at the upper part of the wall as shown in Figure 7.23 (a), (b) and 

(c). The effect of toe fixity is to create a fixed end moment in the wall, reducing the maximum 

bending moment for a given set of conditions but at the expense of increased pile length. The design 

method used (whether Free-end or Fixed-end Method) should also consider the effects of hydrostatic 

pressures and surcharge loads, which are usually added to that due to the soils. 

Deflected shape 

of pile 

(a) (b) (c) 

Figure 7.23 Fixed-end Methpod of Design of Prop Sheet Pile Wall in ranular soils (a) Deflected shape 

of wall (b) Idealized lateral preswsure distribution (c) Simplified Lateral Pressure Distribution 

March 2009 7-29

7.6.3.3 Design of Anchor - General 


In the analysis of anchored steel sheet pile wall, whether using the Fixed-end or Free-end method, 

the tie or strut force, F , per unit length of the wall can be obtained. The restaining anchor must be 

designed to take the required force, F. 

In general, the types of anchor used in sheet pile wall are: 

a) Anchor plates and beams (deadman) Figure 

b) Tie backs 

c) Vertical anchor piles 

d) Anchor beam supported by batter (compression or tension) piles 

These anchors are as shown in Figure 7.24 (a), (b), (c), and (d) respectively. 

(a) Anchor plates and beams 

(b) Tie backs 

(c) Vertical anchor piles 

7-30 March 2009


Figure 7.24 Various types of Anchoring for sheet pile walls (a) Anchor Plate or Beams; (b) Tie Back; 

(c) Vertical Anchor Pile; (d) Anchor Beam with Batter Piles 

The above figures also illustrated the proper locations for placement of various types of anchors. 

Readers can refer to ‘Principles of Getechnical Engineering’ by M. B. Das for further guidance on the 

design of the various types of anchors. 

7.6.3.4 Some Considerations on Sheet Pile Wall Design 

a) Selection of Analysis Method 

Designers must be careful when selecting the design approach to adopt i.e., the Fixed-end or Free 

end method. Walls installed in soft cohesive soils, may not generate sufficient pressure to achieve 

fixity and in those soils it isrecommended that free earth conditions are assumed. Fixed earth 

conditions may be appropriate where the embedment depth of the wall is taken deeper than that 

required to satisfy lateral stability, i.e. to provide an effective groundwater cut-off or adequate 

vertical load bearing capacity. However, where driving to the required depth may be problematic, 

assumption of free earth support conditions will minimise the length of pile to be driven and ensure 

that the theoretical bending moment is not reduced by the assumption of fixity. When designing a 

wall involving a significant retained height and multiple levels of support, the overall pile length will 

often be sufficient to allow the designer to adopt fixed earth conditions for the early excavation 

stages and take advantage of reduced bending moment requirements. 

b) Construction Sequence 

(d) Anchor beam supported by batter (compression or tension) piles 

The design of tied-back or braced system should also consider the sheet pile design requirements at 

each and every stages of the construction sequence, i.e. excavation, strutting, anchoring and 

lowering of ground water table. This construction sequence shall be detailed in the construction 

drawings as wrong construction sequence may cause large changes in the bending moment, shear 

stress and overall stability of the wall. 

c) Permissible Stress of Steel Sheet Pile 

In the design of temporary sheet pile wall, the permissible steel stresses for the structural design of 

the sheet pile can be increased slightly. For instance, Piling Handbook, Archelor Group suggested 

that the permissible steel stresses for temporary works (wall to last not more than 3 months) shown 

in Table 7.3 be used in the structural design in the sheet piles and other steel components of the 

wall such as walins, struts and tie rod. 

March 2009 7-31


d) Design of Cofferdam 

Table 7.4 Permissible Steel Stress of Sheet Pile 

Class of Work Steel grade to EN10248 

S270GP 

(N/mm 2 ) 

S355GP 

(N/mm 2 ) 

Permanent 180 230 

Temporary 200 260 

Cofferdam is a retaining structure, usually temporary in nature, which is used to temporary support 

the sides of deep excavation such as in the construction of multi-level basements and trenches for 

construction of bridge abutment, piers and instalation of deep pipe culverts. Its method of 

construction involved instalation of vertical steel sheet piles to required depth and as excavation 

works progress, a system of wales and struts or prestressed tiebacks (anchors) is installed. 

The earth lateral pressures for the multi-level cofferdam cannot be calculated by the classical 

pressures theories ( Rankine, Coulomb and wedge theories). Readers are advised to refer to 

literatures such as Foundation Design by W.C. Teng or Steel Sheet Piling Design Manual, USS for 

design of this type of wall. 

In addition, the effects of seepage forces and piping need to be considered especially where high 

differential water levels existing between the inner and outer face of the wall. Seepage forces and 

piping or boiling effects can lead to wall instability by reducing passive earth pressure, and in more 

severe cases, can cause liquifaction or ‘quick sand' condition. 

BS8004 1981 provides some guides on the minimum depth of cut-off for cohesionless soils (Table 9, 

pg 47)and shown belows: 

Width, W 

2Y or more 

Y 

0.5Y 

Depth of cut-off, D 

0.4Y 

0.5Y 

0.7Y 

W 

Y 

GWL 

Notes: 

Table 9 ( BS8004 ) 

a) The stability of the wall could 

Idea be increased is to increase by increasing seepage flow the 

path. seepage flow path. 

b) A narrow trench needs a 

Note deeper that cut-off. 

a narrow trench needs a 

c) deeper Value cut-off. 

D obtained to be 

Value compared of D obtained with value to be for 

compared stability. 

with value for stability. 

D 

7-32 March 2009


e) Engineering Software 

Many commercial softwares are also available to facilitate the analysis of retaining wall. Most of 

these software are capable of analyzing more complex and complicated situation e.g. basement 

excavation where high accuracy is required. Some computer programs used the numerical solutions 

to model the soil-structure interaction analysis. Some of these softwares include WALLAP by 

Geosolve, ReWaRD by Geocentrix, FREW by OASYS and many others are available. Finite element 

software such as PLAXIS, SIGMA/W are also becoming increasing more popular as they are able to 

simulate the response of the wall and the soils under various design loadings and construction 

sequence. 

7.6.3.3 Cantilever Steel Sheet Pile Retaining Wall - Example 

A wall is to be built to support a retained height of 3.2m of sandy soils. The effective wall height = 

3.2m + 10% = 3.52m say 3.5m (unplanned excavation allowance is 10% with 0.5m maximum). 

Minimum surcharge loading = 10 kN/m 2 . 

Based on Carquot & Kerisel Chart for K a and K p (Fig. 7.9) 

Loose fine sand K a = 0.3 K p = 0.746 x 6.5 

(Ø = 30°, δ/Ø = -0.5, Reduction Factor for K p = 0.746 – From Fig. 7.9) 

Compact fine sand K a = 0.26 K p = 0.7 x 8.3 = 5.8 

(Reduction Factor for K p = 0.70) 

SURCHARGE 10 kN/m 2 

Overburden kN/m 2 

Active 

Passive 

Water Soil Water Soil 

0.00 10.00 

4.50 m 

6.0 m 1.0 m 

GWL 

Loose Fine Sand 

γ = 17.5 kN/m 3 

γ sat = 19.1 kN/m 3 

= 30° 

Compact Fine Sand 

γ = 18.5 kN/m 3 

γ sat = 19.81 kN/m 3 

= 33° 

γ w = 9.81 kN/m 3 

GWL 

0.30 m Unplanned 3.2 m 

0.00 

0.00 

88.75 

107.25 

0.00 

17.50 

36.00 

0.00 

0.00 

0.00 

-δ/Ø = -0.5 for both soil layer 

58.86 

167.25 

96.00 

58.86 

TYPICAL SECTION 

March 2009 7-33


Note: As ground water levels are the same on both active and passive sides of the wall, pressures 

due to water are ignored. 

Active pressures 

P a at 0.00 m below G.L. in loose sand 

= 0.3 x 10.00 = 3.0 kN/m 2 


= 0.3 x 88.75 = 26.63 kN/m 2 


= 0.260 x 88.75 = 27.89 kN/m 2 


= 0.260 x 167.25 = 43.49 kN/m 2 


= 0.260 x 167.25 = 43.49 kN/m 2 

Passive pressures 

P p at 3.50 m below G.L. in loose sand 

= 4.8 x 0.00 = 0.00 kN/m 2 


= 4.8 x 17.50 + 0.00 = 84.00 kN/m 2 


= 5.8 x 17.50 + 0.00 = 101.50 kN/m 2 


= 5.8 x 36.00 = 208.80 kN/m 2 


= 5.8 x 96.00 = 556.80 kN/m 2 

7-34 March 2009


March 2009 

7-35


Take moments about the toe at 7.022m depth 

Active force 

Force 

(kN/m) 

Moment about toe 

(kNm/m) 

3.0 x 6 = 18.00 x 3.0 = 54.00 

23.63 x 4.5 x 1/2 = 53.17 x 3.00 = 159.50 

20.07 x 1.000 = 20.07 x 1.00 = 20.07 

4.81 x 1.000 x ½ = 2.41 x 0.833 = 2.01 

24.88 x 0.32 = 7.96 x 0.16 = 1.27 

0.83 x 0.32 x ½ = 0.133 x 0.11 = 0.014 

101.74 236.86 

Passive force 

Force 

(kN/m) 

Moment about toe 

(kNm/m) 

84.0 x 1000 x ½ = 42 x 1.65 = 69.30 

101.50 x 1000 = 101.50 x 1.0 = 101.50 

106.80 x 1.000 x ½ = 53.40 x 0.833 = 44.48 

208.30 x 0.50 = 104.15 x 0.167 = 17.36 

29.0 x 0.5 x ½ = 7.25 x 0.167 = 1.21 

308.30 233.85 

Since the passive moment is marginally less than the active moments length is OK. 

To correct the error caused by the use of the simplified method in the depth below the point of 

equal active and passive pressure is increased by 20% to give the pile penetration. 

Let the point of equal pressure be (3.5 + d) below ground level 

Then 84 

23.63 

x d = 3.0 + x (3.5 + d) 

1.00 4.5 

Therefore d = 

18.38 

84 – 5.25 = 0.233m 

Hence the required pile length 

= 3.50 + 0.233 + 1.2 x (2.50 – 0.233) = 6.45m say 6.50m. 

Zero shear occurs at 4.77m below ground level (where the area of the active pressure diagram 

above the level equals the area of the passive pressure diagram above the level). 

Take the moments about and above the level of zero shear (point O): 

kNm/m 

3.0 x 4.77 x ½ x 2.385 = 17.06 

23.63 x 4.5 x ½ x 1.77 = 94.11 

0.056 x 0.27 x ½ x 0.009 = 0.00 

20.08 x 0.27 x 0.135 = 6.73 

-84.00 x 1.000 x ½ x 0.6 = -25.20 

-101.50 x 0.27 x 0.091 = -2.49 

-28.84 x 0.27 x ½ x 0.09 = -0.35 

83.86 

7-36 March 2009


Maximum bending moment = 83.86 kNm/m. 

A partial factor of 1.2 is applied to give the ultimate load. 

Section modulus of pile required 

= 1.2 x 83.36 x 10 3 / 270 = 373 cm 3 /m 

Hence use PU6 piles (z=600 cm 3 /m) not less than 6.50m long in S270GP. 

However the designer will need to check the sustainability of the section for driving and durability. 

March 2009 7-37


REFERENCES 




New York, 1992, 1004 p. 





London 

[6] CGS, “Canadian Foundation Engineering Manual”, (Third edition). Canadian Geotechnical 









[11] DID Malaysia, Retaining Wall 


Kong 

[13] GEO (1993). Guide to Retaining Wall Design (Geoguide 1). (Second edition). Geotechnical 

Engineering Office, Hong Kong, 217 p. 

[14] Harry R.Cedergreen, Seepage, Drainage and Flownet, John Wiley nd Sons. 


New Jersey 



- 494. 

[17] Lambe T.W. and Whitman R.V., "Soil Mechanics", John Wiley 8: Sons, 1969 

[18] McCarthy D.J., "Essentials of Soil Mechanics and Foundations". 


7-38 March 2009



1974. 














March 2009 7-39



7-40 March 2009

CHAPTER 8 GROUND IMPROVEMENT

Chapter 8 GROUND IMPROVEMENT 





8.1 INTRODUCTION .......................................................................................................... 8-1 

8.2 SOIL IMPROVEMENT TECHNIQUES ............................................................................... 8-2 

8.2.1 Removal and Replacement .............................................................................. 8-2 

8.2.2 Surcharging ................................................................................................... 8-3 

8.2.3 SUB SURFACE DRAINAGE IMPROVEMENT SYSTEM ........................................... 8-3 

8.2.3.1 Vertical Drainage System ................................................................. 8-4 

8.2.3.2 Sand Drain System .......................................................................... 8-5 

8.2.3.3 Prefabricated Vertical Drain (PVD) .................................................... 8-5 

8.2.4 Vibro-Floatation ............................................................................................. 8-6 

8.2.4.1 Vibro Compaction ............................................................................ 8-6 

8.2.4.2 Vibro Replacement (Stone Column)................................................... 8-7 

8.2.5 DEEP SOIL MIXING (LIME COLUMN) ................................................................ 8-8 

8.2.5.1 Mix Design ...................................................................................... 8-9 

8.2.6 Dynamic Compaction ...................................................................................... 8-9 

8.2.7 Some Additional Considerations ...................................................................... 8-10 

REFERENCES ....................................................................................................................... 8-12 

APPENDIX 8A: DESIGN OF VERTICAL DRAINAGE SYSTEM ....................................................... 8A-1 

March 2009 8-i




8.1 Typical Properties and Test Standards Specified For Vertical Drain 8-6 



8.1 Distribution of Alluvium Deposits In Peninsular Malaysia 8-1 

8.2 Typical Drainage Directions in Soft Soil During Consolidation Process 8-4 

8.3 Typical Drainage Direction with Vertical Drainage System in Soft Soil during 

Consolidation Process 8-4 

8.4 Typical Schematic Diagram For Vertical Sand Drain System In Embankment 

Construction on Soft Ground 8-5 

8.5 Prefabricated Vertical Drain 8-5 

8.6 Relationships between Particle Size and Available Vibro Techniques 8-6 

8.7 The Schematic Process of Vibro Compaction 8-7 

8.8 Schematic Showing the Installation of Stone Columns (Dry Method) 8-8 

8.9 Mixer Paddle Used In Deep Soil Mixing 8-9 

8.10 Dynamic Compaction 8-10 

8.11 Relationships between U and Tv 8A-2 

8.12 Relationship Of Uh and Tv For Horizontal/Radial Drainage 8A-2 

8.13 Relationship of F(n) and D/dw 8A-4 

8.14 Design Chart for Horizontal Consolidation 8A-5 

8-ii March 2009


8 GROUND IMPROVEMENT 

8.1 

INTRODUCTION 

As land 

becomes scarcer, it is 

often becomes necessary to erect 

structures or buildings on sites 

underlain by poor soils. These sites are potentially troublesome. The most common of these 

problematic soils are the soft saturated clays and silts often found near the mouths of rivers, along 

the perimeter of bays, coast lines and beneath wetlands. 

These soils are very weak and compressible and thus are subjected to bearing capacity and 

settlement problems. They frequently include organic material which further aggravates these 

problems. Areas underlain by these soft soils frequently 

are subject 

to flooding, 

so it often becomes 

necessary to raise 

the ground surface by placing fill. Unfortunately, the weight of these fills 

frequently causes large settlements. 

In Malaysia, deposits of alluvium 

could be found along the coastal line as should in Figure 8.1 which 

illustrated the distribution of alluvial deposits in Peninsular Malaysia. In fact, soft to very soft marine 

clay and silt from a few meters 

to 25 meter depth can be found in 

many areas 

along the coast line 

stretching from Perlis in the north to Johor 

in the south, and also along the coast lines in Sarawak 

and Sabah. 

Figure 8.1 Distribution of Alluvium Deposits In Peninsular Malaysia 

March 2009 

8-1


Fortunately, engineers and contractors have developed methods of coping with these problematic 

soils and have successfully built many large structures on very poor sites. Among the methods used 

(either individually or in combination) include:- 

a) Support the structures on deep foundations that penetrate through the weak soils 

b) Support the structure on shallow foundations and design them to accommodate the weak 

soils 

c) Use a floating foundation, either deep or shallow 

d) Remove the poor material and replace with good materials. This approach is only effective 

if the poor soil material is relatively thin and good replacement soil materials can be easily 

found on site. 

e) Improve the engineering properties of the soils. Various methods of ground improvement 

techniques are available which basically aim to reduce the pore water pressure, reduce 

the volume of voids in the soil, add stronger materials and additives (such as lime or 

cementitious grout) to enhance its soil properties 

f) Avoid the poor ground either by re-alignment or shifting the location of the structures (if 

availability of land is not a constraint) 

The main objectives of ground improvements are to:- 

• Reduce settlement of structures 

• Improve shear strength and bearing capacity of shallow foundations 

• Increase factor of safety against possible slope failure of embankments and dams. 

• Reduce shrinkage and swelling of soils 

The most common techniques often used in our country for solving and stabilizing soft ground 

problems are listed below:- 

a) Structure support system using the shallow foundation or deep foundation and incorporating 

either partially or fully floating foundation principle. Readers are advised to refer to Chapter 5 

and Chapter 9 for shallow foundation and deep foundation respectively. 

b) Soil improvement and stabilization works include 

i) Removal and replacement 

ii) Surcharging 

iii) Sub-surface drainage improvement system 

iv) Vibro floatation 

v) Deep mixing – Lime column 

vi) Dynamic compaction 

8.2 SOIL IMPROVEMENT TECHNIQUES 

8.2.1 Removal and Replacement 

Sometimes poor soils can simply be removed and replaced with good quality compacted fill. This 

alternative is especially attractive if the thickness of the deposit is small, the groundwater table is 

deep and good quality fill material is readily available. If the soil is inorganic and not too wet, then it 

probably is not necessary to haul it away. Such soils can be improved by simply compacting them. In 

this case, the contractor excavates the soil until firm ground is exposed and then places the 

excavated soil back in its original location, compacting it in lifts. This technique is often called 

removed and re-compaction. If necessary, the soil can be reinforced with geosynthetics to spreads 

the applied load over a larger area, thus reducing the change in effective stress and reducing the 

consolidation settlement as well as increasing the bearing capacity. 

8-2 March 2009


Removal and Replacement (or re-compaction) technique is one of the most common and relatively 

less expensive methods used in infrastructures development such as road and earthworks 

construction. However, its usage is limited or constraint by:- 

a. Thickness of unsuitable soft soil 

Often, this technique is only applicable to soft soil layers with thickness less than 3 meter. 

Thick removal may require massive temporary shoring to be in place and end up being 

more costly. 

b. Availability of replacement material 

Availability of replacement material is an important factor as it will govern the overall 

construction cost. Sometimes, light weight material such as Expanded Polystyrene System 

(EPS) is used as an alternative replacement material to minimize excessive consolidation 

settlement and bearing failure of thick fill area. 

8.2.2 Surcharging 

Covering poor soils with a temporary surcharge fill, as shown in Figure 8.3, causes them to 

consolidate more rapidly. When the temporary fill is removed, some or all of the soil is now 

overconsolidated, and thus stronger and less compressible. Often, preloading (by surcharging) has 

been used to improve saturated silts and clays because these soils are most conducive to 

consolidation under static loads. Sandy and gravelly soils respond better to vibratory loads. 

If the soil is saturated, the time required for it to consolidate depends on the ability of the excess 

pore water to move out of the soil voids (see the discussion of consolidation theory in Chapter 4). 

This depends on the thickness of the soil deposit, its coefficient of permeability, and other factors, 

and can be estimated using the principles of soil mechanics. The time required could range from only 

a few weeks to thirty years or more. Allowable construction period is an important factor to 

determine the height of surcharge. Lesser surcharge height will require longer surcharge time. 

For condition where high embankment or surcharge load is required, stage construction can be 

introduced to avoid bearing failure during construction. Consolidation process during stage 

construction will increase soil strength in order to allow higher load at the next stages. 

The consolidation process can be accelerated by an order of magnitude or more by installing vertical 

drains in the natural soil, as discussed in Item 8.2.3. These drains provide a pathway for the excess 

water to escape more easily. Preloading is less expensive than some other soil improvement 

techniques, especially when the surcharge soils can be moved from place to place, thus preloading 

the site in sections. Vertical drains, if needed will increase the cost substantially. 

8.2.3 Sub Surface Drainage Improvement System 

In general sub-drainage system, either horizontal or vertical (or both), can be used to accelerate 

consolidation process by reducing drainage path. These drainage systems provide a pathway for the 

excess water to escape more easily. Vertical drainage system is the most commonly used system for 

embankment constructed on soft soil (provided there are no sand layers or lenses exist in the 

ground) and the directional flows of these drains are as shown in Figure 8.2. The length of the 

drainage path is determined by the thickness of the soft soil or by the existence of any drainage 

layers such as sand layers or lenses. The longer the drainage path, the longer the time required to 

achieve the desired degree of consolidation. 

March 2009 8-3


Figure 8.2 Typical Drainage Directions in Soft Soil During Consolidation Process 

8.2.3.1 Vertical Drainage System 

The introduction of a grid of vertical drains will reduce the traveling distance of the water path 

during consolidation process (refer Figure 8.3), thus increases the rate of consolidation. The 

presence of any natural permeable layers or lenses will further enhance and facilitates horizontal 

water flow toward the vertical drains. This minimizes the excess water pressure generated during 

and after construction and increases the rate of settlement. 

Generally there are 2 common vertical drainage systems available in the market, namely:- 

a) Sand drain system 

b) Prefabricated vertical drain (PVD) system 

Figure 8.3 Typical Drainage Direction with Vertical Drainage System In Soft Soil During Consolidation 

Process 

8-4 March 2009


8.2.3.22 Sand Drain System 

Sand drain system 

has been introduced since 1930s as the vertical drainage techniquee for soft 

ground. In general sand column 

are installed in grid pattern with spacing ranges from 2 – 3m center 

to center. The common diameter adopted ranges from 200mm to 400mm and the allowable 

depth of 

treatment can be as deep as 30m. One of the typical examples of sand drain application is the 

manmade island for the Kansai Airport Japan in 1990s. The application of sand drain has slowly been 

replaced by Prefabricated Vertical Drain due mainly to its speed, ease of construction and relatively 

cheaper cost. 

Figure 

8.4 Typical Schematic Diagram For Vertical Sand 

Drain System In Embankment Construction 

on Soft Ground 

8.2.3.33 Prefabricated Vertical Drain (PVD) 

PVD has been widely used as 

vertical drainage system. It is a manufactured drain made from 

synthetic material. In general, PVD is very thin material, approximately 4mm with a common width 

of 100mm. The very thin material would minimize clay 

smearing during installation whichh reduces 

the efficiency of the 

drain. PVD is slowly replacing the use of sand drain because of the cheaper cost 

and fast installation. Figure 8.6 shows a picture of a typical PVD available in the market. 

Figure 8.5 Prefabricated Vertical Drain 

March 2009 

8-5


PVD normally consists of 2 main components, i.e., the center core and the filtering jacket. The drain 

cores are of flexible type which allows freee flow of water along and/or acrosss the drain core. The 

filter is of the non-woven geo-fabric type with specific pore size distribution. The 

drain core and filter 

are made of one or combination of the following materials: polyester, polyamide, polypropylene, 

polyethylene or any 

other natural polymeric 

material. 

The filtering jacket acts as a natural soil filter surface which inhibit 

movement of soil particles while 

allowing 

passage of water into the drain. Thus, it acts as the exterior surfaces and prevents closure 

of the internal drain 

flow paths under lateral soil pressures. 

The PVD center core serves to provide the internal flow 

paths along 

the drain and at the same time, 

provide 

support to the filter jacket to maintain the drain configuration and shape. It also provides 

some resistance to longitudinal stretching as well as buckling of the drain. 

Reader 

can refer to Appendix 

system. 

8A for a more detail discussion on 

the design 

of vertical 

drainage 

8.2.4 

Vibro-Floatation 

The process of improving loose 

granular ground soil with depth vibrators started in the 1930s. With 

the advancement of technology, vibro-floatation technique has also been used to treat cohesive soil. 

Vibro-floatation can 

be dividedd into two main categories, namely; Vibro Compaction and Vibro 

Replacement. Vibro 

Compaction basically is 

used to treat granular soils by densifying loose 

granular 

soils by 

means of depth vibrator. As for Vibro Replacement, it is 

used to treat cohesivee soils by 

partially 

replacing the cohesive soils with granular soils (in this case, vibro replacement is sometimes 

referred 

to as stone column) ). Figure 8.6 

shows the relationship between soil types and the 

appropriate method 

of vibro floatation. 

Figure 8.6 Relationships between Particle Size and Available Vibro Techniques 

8.2.4.1 

Vibro 

Compaction 

The principle behind this method is that the cohesiveless soil i.e., sand and gravel can be densified 

by means of vibration. The vibratory action 

of the depth vibrator is used to temporarily reduce the 

particular friction between the particles and 

rearrange soil particles 

in a denser state. The effect of 

vibro densification 

can increasee the shear strength of the existing ground and reduce the total and 

differential settlement. 

8-6 

March 2009


The vibrator penetrates the soil by means of water jets and once at full depth, it is gradually 

withdrawn leaving behind a column of well compacted soil. Figure 8.7 illustrated the schematic 

processs of vibro compaction. 

To achieve a mass densification, the entire area is compacted by 

column 

points in a triangle or square pattern. This technique is well suited for the densification of 

relatively clean (fines content up to about 10 to 15%) 

granular soils such as sands and gravels. A 

major benefit of this method is that no additional materials are necessary which makes it a very 

economical technique. The extent and effectiveness of 

the techniques in improving the compaction 

of the soil can be determined easily by sounding tests such as cone penetration test or electric 

piezocone. 

Figure 8.7 The Schematic Process of Vibro Compaction 

8.2.4.22 Vibro 

Replacement (Stone Column) 

Vibro replacement 

is a technique used to improve sandy soils with high fines contents (>15%) and 

cohesive soils such as silts and clays. In this method columns made 

up of stones are installed in the 

soft ground using the depth vibrator. The vibrator is used to first create a hole in 

the ground 

which is 

then filled with stones as the vibrator is withdrawn. The stones are 

then laterally displacedd into the 

soil by 

subsequent 

re-penetration of the vibrator. In 

this manner a column made up 

of well 

compacted stone fill with diameters typically ranging between 0.7 

m and 1.1 m is installed in the 

ground. 

Two methods of installation namely the ‘wet’ and ‘dry’ methods are 

used for installation of the stone 

columns. In the wet method, water jets are 

used to create the hole and to assist in penetration. In 

the dry 

method, the hole is created by the vibratory energy and induced pulll down force. Typical 

installation process in the case of dry method is schematically shown in Figure 8.8. This technique of 

soil improvement can be used for nearly all types of soils. 

March 2009 

8-7


Figure 8.8 Schematic Showing the Installation of Stone Columns (Dry 

Method) 

The Vibro Replacement technique provides an economical and flexible solution, which can readily be 

adaptedd to varying ground conditions. Vibro 

Replacement techniquee can improved the soil conditions 

in various ways, among which are: 

• Compaction of the subsoil and increase in density 

• Improvement in the 

stiffness of 

the subsoil to decrease excessive settlement 

• Improvement in the 

shear strength of the subsoil to decrease the risk of failure 

• Increase in the mass of the subsoil to mitigate ground vibrations 

• Ability to carry very 

high loads since columns are highly ductile 

• Rapid consolidationn of the subsoil 

Stone column improvement shall not be treated as structural solution. Dense stone columns 

installed 

and the surrounding soil is considered as a composite matrix. Shear strength consider after 

treatment is not limited to stone column but subjected to overall strength increase. Overall 

composite strength shall be considered in stability design. The common design approach adopted in 

stone column is using Priebe’s 

method which developed by Heinz J. Priebe 1995 from Keller. In 

Priebe’s 

method, improvement 

factors are 

calculated to be column spacing, diameter, constraint 

modulus and etc. The common 

diameter of stone column adoptedd in Malaysia ranges from 

900mm 

to 1200mm diameter. Depth of 

treatment is subjected to loading, soil stratum, 

need for settlement 

/stability. 

Testing 

of the soil improvement, after installation of the stone columns in coarse-grained soils is 

usually performed with either static or dynamic penetrometer tests (CPT or DPT). However for stone 

columns constructed in fine-grained soils it is common practice to carry out load 

tests directly on the 

columns. 

8.2.5 

Deep Soil Mixing 

(Lime Column) 

Deep soil mixing (DSM) technology is a development of 

the lime-cement column method, which was 

introduced almost 30 years ago. It is a form of soil improvement involving the introduction and 

mechanical mixing of in-situ soft and weak soils with a cementitious 

compound such as lime, cement 

or a combination of both in different proportions. The mixing of the cementitious compound is 

facilitated with a rotary paddle as shown in Figure 8.9. The mixture 

is often referred to as the 

8-8 

March 2009


binder. The binder is injected into the soil in a dry form. The moisture in the soil is utilized for the 

binding process, resulting in an improved soil with higher shear strength and lower compressibility. 

The removal of the moisture from the soil also results in an improvement in the soft soil surrounding 

the mixed soil. 

Figure 8.9 Mixer Paddle Used In Deep Soil Mixing 

Typical applications of the deep soil mixing method include foundations of embankment fill for 

highway and railway, slope stabilization, stabilization of deep excavation and foundations for housing 

development. The anticipated amounts of binding agents commonly used are approximately 100 – 

150 kg/m 3 in silty clay and clayey silt materials. The strength develops differently over time 

depending on the type of soil, amount of binder and proportion used. In most cases, the strength 

starts to increase after a few hours and then continues to increase rapidly during the first week. In 

normal cases, approximately 90% of the final strength is reached after about three weeks. 

8.2.5.1 Mix Design 

Detailed site investigation and laboratory tests are required to determine the optimum lime content 

for soil stabilization. In general, lime stabilization is suitable for ground with low sulphide and organic 

content. It is also effective for silty ground with low plasticity. The optimum lime percentage is 

approximately 3% but increases with water content. However if lime content exceeded the optimum 

content, shear strength of treated ground will be reduced. The increase in the shear strength after 

improvement varies, and ranges from 5-10 kPa to 100kPa. Generally shear strength increment 

reduces with increment of liquid limit. 

The soil strength increase gradually through the pozzolonic reaction between lime, aluminate and 

silicate in the soil (clay). The percentage of clay shall be more than 20%. For normal case, the 

mixture of silt and clay shall be greater than 35% and plasticity shall be greater than 10%. If the 

percentage of clay does not fulfill the condition above, cement and fly ash shall be added. 

For soil improvement using lime mixing in organic soil, shear strength increment is rather small. 

Usually, gypsum is added to unslaked lime to stabilize the organic soil. The mixture is of 

approximately ¼ to ½ of gypsum to ¾ ~ ½ unslaked lime. 

8.2.6 Dynamic Compaction 

Dynamic compaction consists of using a heavy tamper that is repeatedly raised and dropped with a 

single cable from varyingn heights to impact the ground. The mass of the tampers generally ranges 

from 20 tonnes to 200 tonnes and drop height range from 20 to 40m. The energy is generally 

applied in phases on a grid pattern over the entire area using single or multiple passes. Following 

March 2009 8-9


each pass, the craters are either levelled with a dozer or filled with 

granular fill material before the 

next pass of energy 

is applied. Figure 8.10 shows the schematic of the dynamic compaction process. 

Figure 8.10 Dynamic Compaction 

All of the energy is 

applied from 

existing grade and the 

degree of improvement is a function of the 

energy applied i.e., 

the mass of the tamper, the drop height, the grid spacing 

and the number of 

drops at each grid point location. 

The application of 

dynamic compaction shall take into consideration the noise and vibration 

disturbances to the surrounding. Excessive vibration 

may cause 

distresses to the neigbouring 

structures. 

In situ test such as SPT, CPT or Piezocone can be used during and after completion of dynamic 

compaction to verify whether the desiredd improvement has not 

been achieved. If necessary, 

additional energy could be applied to further improve 

the densification and improvement of the 

ground. 

8.2.7 

Some 

Additional 

Considerations 

a) The selection of ground improvement methods is subjected to the following 

criterions:- 

i) Cost effectiveness of the treatment method as 

compared to the overall project cost 

ii) The availability of the 

treatment method in the country 

iii) 

Types of soil to be treated 

iv) 

Long term 

and differential settlement requirements for the 

structures 

b) The 

construction rate of the earthworks is usually faster than the dissipation of pore water 

pressure (especially in low permeability 

clay soil). The initially high excess pore water pressure 

developed in the ground due to rapid construction will reduce the effectivee strength of the soil 

and 

may lead to ground instability. However, the excess pore pressure will slowly dissipate with 

time, thus increases the effective stresss of the soil which eventually increases the stability of the 

ground. Hence, 

total stresss analysis with undrainedd condition, which is usually the most critical 

condition, is used in the design of ground treatment. 

8-10 

March 2009


c) Soils subjected to improvement works are usually very soft in nature. Standard Penetration Test 

(SPT) is not suitable for soft soil layer. It is advisable to retrieve undisturbed soil samples from 

the ground for laboratory tests which include Undrained Unconsolidated (UU) Triaxial test and 

One Dimensional Consolidation Test using Odeometer. In addition, in-situ tests such as Vane 

Shear test and Piezocone are recommended in soft soils sensitive to disturbance such as marine 

clay is highly recommended. 

d) Transition zone shall be provided in the ground improvement design if the project used more 

than one type of ground improvement methods. This is most crucial if the ground improvement 

methods pose a different allowable long term settlement, e.g., bridge and bridge approach, 

culverts etc. 

e) Due to the complexities and uncertainties of the ground conditions as well as the simplification of 

design formulae in the analysis and design, it is strongly recommended that the instrumentation 

monitoring scheme shall be provided during the construction works for design verification 

purposes. Some provisions in the Bill of Quantities shall also be provided to cater for any design 

changes during construction. 

March 2009 8-11


REFERENCES 

[1] ASCE (1987). Soil Improvement – A ten Year Update, Geotechnical Special Publication No. 

12, edited by J.P. Welsh. 

[2] Bowles, J.E. (1988). Foundation Analysis and Design, 4 th ed., McGraw-Hill, New York. 

[3] Broms, B.B. (1993). Lime Stabilization. “Chapter 4 in Ground Improvement, edited by M.P. 

Moseley, CRC Press, Boca Raton, Florida, pp. 65-99. 

[4] Broms, B.B., and Forssblad, L. (1969). “Vibratory Compaction of Cohesionless Soils. 

“Proceedings of the Seventh International Conference on Soil Mechanics and Foundation 

Engineering, Specialty Session No. 2, pp. 101-118. 

[5] Broomhead, D., and Jasperse, B.H. (1992). “Shallow Soil Mixing- a Case History. “Grouting, 

Soil Improvement and Geosynthetic, edited by R.H. Borden, R.D. Holtz, and I. Juran, ASCE 

Geotechnical Special Publication no. 3o, vol. 1, pp. 564 – 576. 


[7] Coduto, D. P., (2001) Foundation Design – Principles and Practices, Prentice Hill Inc. 

[8] Das, B.M. (1983). Advanced Soil Mechanics, Hemisphere Publishing, New York. 

[9] Dept. of the Navy, Bureau of Yards and Docks, Washington D.C.,NAVFAC DM.-7.3, April 

1983, Soil Dynamics, Deep Stabilization and Special Geotechnical Construction 

[10] Duncan, J.M. & Poulos, H.G. (1981). Modern techniques for the analysis of eng 

[11] ineering problems in soft clay. Soft Clay Engineering, Elsevier, New York, pp 317-414. 



[13] FHWA (1979). Soil Stabilization in Pavement Structures- a User’s Manual, Report no. FHWA- 

IP-80-2, Federal Highway Administration, Washington, D.C., October. 

[14] Hausmann, M.R. (1990). Engineering Principles of Ground Modification, McGraw-Hill, New 

York. 

[15] Koerner R.M . Construction and Geotechnical Method in Foundation Engineering, McGraw 

Hill, 1985. 


[17] Mesri G., discussion of New Design Procedure for stability of Soft Clays. by Charles C. Ladd 


Paper 10664. April 1975. pp. 409 - 412. 

[18] Nayak N. V. I II, Foundation Design Manual. Dhanpat Rai a Sons I 1982. 


1974. 

8-12 March 2009


[20] O.G., and Metcalf, J.B. (1973), Soil Stabilization: Principles and Practice, John Wiley & Sons, 

New Ingles. 

[21] PCA(1979). Soil-Cement Construction handbook, Portland Cement Association, Skokie, 

Illinois. 

[22] Sherwood, P.T.(1962). Effect of Sulfates on Cement-and Lime-Stabilized Soils. Highway 

Research Board Buletin No. 353: Stabilization of Soils with Portland Cement, Washington, D.C., pp. 

98-107. Also in Roads and Road Construction, vol. 40, February, pp. 34-40. 

[23] Sokolovich, V.E., and Semkin, V.V. (1984), Chemical Stabilization of Loess Soils. Soil 

Mechanics and Foundation Engineering, vol. 21, no. 4, July-August, pp. 8-11. 




[26] Thomson, M.R. (1966). Shear Strength and Elastic Properties of Lime-Soil Mixtures. 

Highway Research Record No. 139: Behaviour Characteristics of Lime-Soil Mixtures, highway 

Research Board, Washington, D.C., pp. 1-14. 

[27] Thonson, M.R. (1969). Engineering Properties of Soil-Mistures. Journal of Materials, ASTM, 

vol. 4, no. 4, December. 

[28] TRB (1987). Lime Stabilization: Reactions, Properties, Design, and Construction, State of the 

Art Report 5, Transportation Research Board, Washington, D.C. 

March 2009 8-13



8-14 March 2009


APPENDIX 8A: DESIGN OF VERTICAL DRAINAGE SYSTEM 

The principal objective of soil pre consolidation, with or without PVD, is to achieve a desired degree 

of consolidation within a specified period of time. The design of pre consolidation with PVDs requires 

the evaluation of drain and soil properties (both separately and as a system) as well as the effects of 

installation. 

For one dimensional consolidation with drains, only consolidation due to one dimensional (vertical) 

seepage to natural drainage boundaries is considered. The degree of consolidation can be measured 

by the ration of the settlement at any time to the total primary settlement that will (or is expected 

to) occur. This ratio is referred to as Ū, the average degree of consolidation. 

By definition, one dimensional consolidation is considered to result from vertical drainage only, but 

consolidation theory can be applied to horizontal or radial drainage as well. Depending on the 

boundary conditions consolidation may occur due to concurrent vertical and horizontal drainage. The 

average degree of consolidation, Ū, can be calculated from the vertical, horizontal or combined 

drainage depending on the situation considered. 

With Vertical drains the overall average degree of consolidation, Ū, is the result of the combined 

effects of the horizontal (radial) and vertical drainage. The combined effect is given by:- 

Ū = 1 – ( 1 – Ū h ) (1 – Ū v ) (8.1) 

where, 

Ū = overall average degree of consolidation 

Ū h = average degree of consolidation due to horizontal (radial) Drainage 

Ū v = average degree of consolidation due to vertical drainage. 

The graph of Ū vs log time for both the vertical and horizontal drainage in shown in Figure 8.11 and 

Figure 8.12 respectively. 

March 2009 8A-1


Figure 8.11 Relationships between U and Tv 

Figure 8.12 Relationship Of Uh and Tv For Horizontal/Radial Drainage 

The design of PVD system requires the prediction of the rate of dissipation of excess pore pressures 

by radial seepage to 

vertical drains as well as evaluating 

the contribution of vertical drainage. 

The first comprehensive treatment of the radial drainage problem 

was presented by Barron who 

studiedd the theory of vertical sand drains. Barron works was based on simplifying assumptions of 

Terzaghi’s one-dimensional linear consolidation theory. The most widely used simplified solution from 

Baron’s 

analysis provides the relationship of time, drain diameter, spacing, coefficient of 

consolidation and the average degree of consolidation. 

8A-2 

March 2009


t = (D 2 /8C h ) F(n) ln (1/(1- Ū h )) (8.2) 

where, 

t = time required to achieve Ū h 

Ū = average degree of consolidation due to horizontal drainage. 

D = diameter of the cylinder of influence of the drain (drain influence zone) 

C h = coefficient of consolidation for horizontal drainage 

F(n) = Drain spacing factor 

= ln (D/d) – ¾ 

D = diameter of a circular drain 

Equation 8.2 was further modified by Hasbo to be applied to band-shape PVD and to include 

consideration of disturbance and drain resistance effects. 

t = (D 2 /8C h ) (F(n) + Fs + Fr) ln (1/(1- Ū h )) (8.3) 

where, 

t = time required to achieve Ū h 

Ū = average degree of consolidation at depth z du to horizontal drainage 

D = diameter of the cylinder of influence of the drain (drain influence zone) 

C h = coefficient of consolidation for horizontal drainage 

F(n) = Drain spacing factor 

= ln (D/d w ) – ¾ 

D = diameter of a circular drain 

d w = equivalent diameter 

Fs = factor for soil disturbance 

= ((k h /k s ) – 1) ln (d s /d w ) 

k h = the coefficient of permeability in the horizontal direction in the undisturbed soil 

k s = the coefficient of permeability in the horizontal direction in the disturbed soil 

d s = diameter of the idealized disturbed zone around the drain 

Fr = factor for drain resistance 

= πz (l – z) (k h /q w ) 

z = distance below top surface of the compressible soil later 

L = effective drain length; length of drain when drainage occurs at one end only; half 

length of drain when drainage occurs at both ends 

q w = discharge capacity of the drain (at gradient = 1.0) 

Equation 8.3 can be simplified to the ideal case by ignoring the effect of soil disturbance and drain 

resistance (Fs and Fr = 0) the resulting ideal case equation is equivalent to Barron’s solution: 

t = (D 2 /8C h ) F(n) ln (1/(1- Ū h )) (8.4) 

Therefore, in the ideal case, the time for a specified degree of consolidation simplifies to be a 

function of soil properties (C h ), design requirement (Ū h ) and design variables (D, d w ). 

March 2009 8A-3


Figure 8.13 Relationship of F(n) and D/dw 

Figure 8.13 shows the relationship of F(n) to D/d w for the ideal case. Within a typical range of D/d w , 

F(n) ranges from approximately 2 to 3. 

The theory of consolidation with radial drainage assumes that the soil is drained by a vertical drain 

with circular section. The radial consolidation equations include the drain diameter, d. A band shape 

PVD drain must therefore be assigned as “equivalent diameter”, d w. For design purposes, it is 

reasonable to calculate the equivalent diameter as:- 

where, 

d w = (2(a+b)/π) (8.5) 

a 

b 

= width of the band – shaped drain cross section 

= thickness of a band-shaped drain cross section 

Equation A8.5 can be further simplified to 

d w = (a + b) /2 (8.6) 

8A-4 March 2009


% of Consolidation 

Consolidationn period (month) 

Spacing (m) 

C h m 2 /year 

Figure 8.14 Design Chart for Horizontal Consolidation 

The Design Chart shown in Figure 8.14 can 

be used as a preliminary guide for PVD design. Simple 

input parameter such as drain spacing, degree of consolidation, required consolidation duration and 

coefficient of horizontal consolidation are used for PVD design. 

In context of local Malaysian soft soil, the typical spacing of PVD ranges from 

1.0 to 1.5m 

c/c. In 

some construction, to further reduce the consolidation period, additional surcharge load is used. 

Some of the typical properties specified for Prefabricated Vertical Drain (PVD) are as shown 

in Table 

8.1 below. The actual limiting values of the 

properties can be obtained from the 

various suppliers or 

manufacturers: 

March 2009 

8A-5


Table 8.1 Typical Properties and Test Standards Specified For Vertical Drain 

Criteria Properties Standard 

General Thickness ASTM D5199 

Constructability Tensile Strength (dry and Wet) 

Grab 

Strip 

Wide Width 

ASTM D4132 

ASTM D1182 

ASTM D5035 

Tear Strength 

ASTM D4533 

Puncture resistance 

ASTM D4833 

Abrasion resistance 

ASTM D4881 

Ultra violet stability 

ASTM D4355 

Hydraulic Permeability / permittivity ASTM D4491 

Apparent opening size (O 95 

) 

ASTM D4751 

Discharge capacity 

ASTM D4711 

8A-6 March 2009

CHAPTER 9 FOUNDATION ENGINEERING

Chapter 9 FOUNDATION ENGINEERING 



List of Tables ...................................................................................................................... 9-iii 

List of Figures ..................................................................................................................... 9-iii 

9.1 INTRODUCTION .......................................................................................................... 9-1 

9.2 DEEP FOUNDATION ..................................................................................................... 9-2 

9.2.1 General ......................................................................................................... 9-2 

9.2.2 Classification of Piles ....................................................................................... 9-2 

9.2.2.1 Precast Reinforced Concrete Piles ....................................................... 9-2 

9.2.3 Pile Foundation Design .................................................................................... 9-6 

9.2.3.1 General ............................................................................................ 9-6 

9.2.3.2 Design Philosophies ........................................................................... 9-6 

9.2.3.4 Pile Capacity ..................................................................................... 9-8 

9.2.4 Pile Loading Tests ........................................................................................ 9-13 

9.2.4.1 General .......................................................................................... 9-13 

9.2.4.2 Timing of Pile Tests ......................................................................... 9-14 

9.2.4.3 Static Pile Loading Tests .................................................................. 9-14 

9.2.5 Equipment ................................................................................................... 9-17 

9.2.5.1 Measurement of Load ...................................................................... 9-17 

9.2.5.2 Measurement of Pile Head Movement ............................................... 9-19 

9.2.5.3 Test Procedures .............................................................................. 9-21 

9.2.5.4 Instrumentation .............................................................................. 9-24 

9.2.5.5 Interpretation of Test Results ........................................................... 9-25 

9.2.6 Dynamic Loading Tests ................................................................................. 9-27 

9.2.6.1 General .......................................................................................... 9-27 

9.2.6.2 Test Methods .................................................................................. 9-27 

9.2.6.3 Methods of Interpretation ................................................................ 9-28 

9.2.6.4 Recommendations on the Use of Dynamic Loading Tests .................... 9-29 

9.3 LATERALLY LOADED PILES ......................................................................................... 9-29 

9.3.1 Introduction ................................................................................................. 9-29 

9.3.2 Lateral Load Capacity of Pile .......................................................................... 9-31 

9.3.3 Inclined Loads .............................................................................................. 9-39 

9.3.4 Raking Piles in Soil ........................................................................................ 9-39 

9.3.5 Lateral Loading ............................................................................................ 9-40 

March 2009 9-i


9.3.5.1 General .......................................................................................... 9-40 

9.3.5.2 Equivalent Cantilever Method ........................................................... 9-41 

9.3.5.3 Subgrade Reaction Method .............................................................. 9-41 

9.3.5.4 Elastic Continuum Method ................................................................ 9-43 

9.4 PILE GROUP .............................................................................................................. 9-45 

9.4.1 General ....................................................................................................... 9-45 

9.4.2 Minimum Spacing of Piles ............................................................................. 9-46 

9.4.3 Ultimate Capacity of Pile Groups .................................................................... 9-46 

REFERENCES ..................................................................................................................... 9-48 

9-ii March 2009




9.1 Advantages and Disadvantages of Machine-dug Piles 9-4 

9.2 Advantages and Disadvantages of Hand-dug Caissons 9-5 

9.6 Tolerance of Installed Piles 9-46 



9.1 Types of Foundation 9-1 

9.2 Estimation of Negative Skin Friction by Effective Stress Method 9-13 

9.3 Typical Arrangement of a Compression Test using Kentledge 9-15 

9.4 Typical Arrangement of a Compression Test using Tension Piles 9-16 

9.6 Typical Instrumentation Scheme for a Vertical Pile Loading Test 9-21 

9.7 Typical Load Settlement Curves for Pile Loading Tests (Tomlinson, 1994) 9-26 

9.8 Failure Modes of Vertical Piles under Lateral Loads (Broms, 1914a) 9-30 

9.9 Coefficients Kqz and Kcz at Depth z for Short Piles Subject to Lateral Load 

(Brinch Hansen, 1911) 9-33 

9.10 Ultimate Lateral Resistance of Short Piles in Granular Soils (Broms, 1914a) 9-34 

9.11 Ultimate Lateral Resistance of Long Piles in Granular Soils (Broms, 1914b) 9-35 

9.12 Influence Coefficients for Piles with Applied Lateral Load and Moment (Flexible 

Cap or Hinged End Conditions) (Matlock & Reese, 1910) 9-37 

9.13 Influence Coefficients for Piles with Applied Lateral Load (Fixed against Rotation at 

Ground Surface) (Matlock & Reese, 1910) 9-38 

9.14 Analysis of Behaviour of a Laterally Loaded Pile Using the Elastic Continuum 

Method (Randolph, 1981a) 9-44 

March 2009 

9-iii



9-iv March 2009


9 DEEP FOUNDATION ENGINEERING 

9.1 

INTRODUCTION 

In general, deep foundation using piles are relied upon to transfer the load 

acting on the 

superstructures in situations where the use of shallow foundations becomes inadequate or unreliable. 

Some of the situationss where piles are required are as follows: 

• To 

transfer loads through water or soft soil to a suitable bearing stratum by means of end 

bearing of the piles (end bearing or point bearing piles). 

• To 

transfer loads to a depth of a relatively 

weak soil by 

means of "skin friction" along the length 

of the piles (friction piles). 

• To 

compact granular soils, thus increasing 

their bearing capacity (compaction piles). 

• To 

carry the foundation through the depth of scour to provide safety in the event the soil is 

eroded away. 

• To 

anchor down 

the structures subjected to uplift due to hydrostatic (Pressure or overturning 

moment (tension pile or uplift pile). 

• To 

provide anchorage against horizontal pull from sheetpiling walls or other pulling forces 

(anchor piles). 

• To 

protect water front structures against impact from ships or other floating objects (fender piles 

and dolphins). 

• To 

resist large horizontal or inclined forces (batter piles). 

Foundation can be divided into two main categories, namely shallow foundation and deep foundation. 

The common type of foundation is shown in Figure 9.1 below. 

Foundations 

Shallow 

Foundations 

Deep 

Foundations 

Spread 

Footings 

Mat 

Foundations 

Driven 

Piles 

Drilled 

Shafts 

Auger Cast 

Piles 

Figure 9.1 Types of Foundation 

This Chapter discusses the principles and design of deep foundation. For shallow foundation, reader 

can refer to Chapter 4 and Chapter 5 for more detailed discussion on 

soil settlement and bearing 

capacity theory respectively. 

March 2009 

9-1


9.2 DEEP FOUNDATION 

9.2.1 General 

Deep foundation is usually used when tructural load is relatively high and/or the ground condition does 

not allow for shallow foundation system. Sometimes due to high load, required spread footing are too 

large and not economical. For some special structures, i.e., bridge pier, dock etc, pile foundation is 

adopted because the foundation is subjected to scour or undermining. Generally deep foundation 

system is also preferable where the structures are subjected to high uplift force or lateral force. 

9.2.2 Classification of Piles 

There are many types of pile classification adopted. In general, piles can be classified according to:- 

a) The type of material forming the piles, 

b) The mode of load transfer, 

c) The degree of ground displacement during pile installation and 

d) The method of installation. 

Pile classification in accordance with material type (e.g. steel and concrete) has drawbacks because 

composite piles are available. A classification system based on the mode of load transfer will be 

difficult to set up because the proportion of shaft resistance and end-bearing resistance that occurs in 

practice usually cannot be reliably predicted. 

In the installation of piles, either displacement or replacement of the ground will predominate. A 

classification system based on the degree of ground displacement during pile installation, such as that 

recommended in BS 8004 (BSI, 1981) encompasses all types of piles and reflects the fundamental 

effect of pile construction on the ground which in turn will have a pronounced influence on pile 

performance. Such a classification system is therefore considered to be the most appropriate. In this 

document, piles are classified into the following four types: 

(a) 

(b) 

(c) 

(d) 

Large-displacement piles, which include all solid piles, including precast concrete piles, and steel 

or concrete tubes closed at the lower end by a driving shoe or a plug, i.e. cast-in-place piles, 

large diameter spun pile etc. 

Small-displacement piles, which include rolled steel sections such as H-piles and open-ended 

tubular piles. However, these piles will effectively become large-displacement piles if a soil plug 

forms. 

Replacement piles, which are formed by machine boring, grabbing or hand-digging. The 

excavation may need to be supported by bentonite slurry, or lined with a casing that is either left 

in place or extracted during concreting for re-use. 

Special piles, which are particular pile types or variants of existing pile types introduced from 

time to time to improve efficiency or overcome problems related to special ground conditions. 

9.2.2.1 Precast Reinforced Concrete Piles 

Precast reinforced concrete piles are common nowadays in Malaysia. These piles are commonly in 

square sections ranging from about 250 mm to about 450 mm with a standard length varies from 1m 

to 12m. The lengths of pile sections are often dictated by the practical considerations including 

transportability, handling problems in sites of restricted area and facilities of the casting yard In 

general, and the maximum allowable axial loads is subjected to the structural capacity designed by the 

manufacturer and it can be up to about 1 000kN. These piles can be lengthened by coupling together 

during installation. Joining method commonly adopted in Malaysia is using wielding of the end plate of 

the piles. 

9-2 March 2009


This type of pile is not suitable for driving into ground that contains a significant amount of boulders 

or corestones and very hard sand lenses. 

i) Precast Prestressed Spun Piles 

Precast prestressed spun concrete piles used in Malaysia are closed-ended tubular sections of 400 mm 

to 1000 mm diameter with maximum allowable axial loads up to about 3000 kN. Special large diameter 

spun piles with diameter greater than 1000mm are also available but the demand is low. Pile sections 

are normally 12 m long and are usually welded together using steel end plates. 

Precast prestressed spun concrete piles require high-strength concrete and careful tight QA/QC control 

during manufacture. Casting is usually carried out in a factory where the curing conditions can be 

strictly regulated. Special manufacturing processes such as compaction by spinning or autoclave curing 

can be adopted to produce high strength concrete up to about 75 MPa. Such piles may be handled 

more easily than precast reinforced concrete piles without damage. Steam curing is usually adopted in 

the casting yard to shorten casting time and to ensure the quality of the pile. 

ii) Small-Displacement Piles 

Small-displacement piles are either solid (e.g. steel H-piles) or hollow (open-ended tubular piles, i.e., GI 

pipes) with a relatively low cross-sectional area. This type of pile is usually installed by percussion 

method. However, a soil plug may be formed during driving, particularly with tubular piles, and 

periodic drilling out may be necessary to reduce the driving resistance. A soil plug can create a greater 

driving resistance than a closed end, because of damping on the inner-side of the pile. 

Bakau pile is considered to be a small displacement pile. However, due to the conservation of the 

mangrove forest and the coastal line of Malaysia. Bakau piles are not allowed to be used special permit 

is required if imported bakau pile is used. 

iii) Replacement Piles 

Replacement or bored piles are mostly formed by machine excavation. When constructed in condition 

with high ground water table, the pile bore will need to be supported using steel casings, concrete rings 

or drilling fluids such as bentonite slurry, polymer mud, etc to avoid collapsing of drilled hole. 

Excavation of the pile bore may also be carried out by hand-digging in the dry; and the technique 

developed in Hong Kong involving manual excavation is known locally as hand-dug caissons. 

Machine-dug piles are formed by rotary boring, or percussive methods of boring, and subsequently 

filling the hole with concrete. Piles with 100 mm or less in diameter are commonly known as smalldiameter 

piles. Piles greater than 1000 mm diameter are referred to as large-diameter piles. 

a) Machine Bored Piles 

The advantages and disadvantages of machine-dug piles are summarized in Table 9.1. 

March 2009 9-3


Table 9.1 Advantages and Disadvantages of Machine-dug Piles 

Advantages 

i. No risk of ground heave induced by pile 

driving. 

ii. Length can be readily varied. 

ii. Spoil can be inspected and compared with 

site investigation data. 

v. Structural capacity is not dependent on 

handling or driving conditions. 

v. Can be installed with less noise and vibration 

compared to displacement piles. 

vi. Can be installed to great depths. 

vii. Can readily overcome underground 

obstructions at depths. 

Disadvantages 

a. Risk of loosening of sandy or gravelly soils 

during pile excavation, reducing bearing 

capacity and causing ground loss and hence 

settlement. 

b. Susceptible to bulging or necking during 

concreting in unstable ground. 

c. Quality of concrete cannot be inspected after 

completion except by coring. 

d. Unset concrete may be damaged by 

significant water flow. 

e. Excavated material requires disposal, the 

cost of which will be high if it is 

contaminated. 

f. Base cleanliness may be difficult to achieve, 

reducing end-bearing resistance of the piles. 

b) Mini / Micro Bored Piles 

Mini-piles generally have a diameter between 100 mm and 400 mm. One or more high yield steel bars 

are provided in the piles. In Malaysia, used high yield steel pipes are commonly used as the 

reinforcement for micro piles. 

Construction can be carried out typically to about 10 m depth or more, although verticality control will 

become more difficult at greater depths. Mini-piles are usually formed by drilling rigs with the use of 

down-the-hole hammers or rotary percussive drills. They can be used for sites with difficult access or 

limited headroom and for underpinning. In general, they can overcome large or numerous obstructions 

in the ground. 

Mini-piles are usually embedded in rock sockets. Given the small-diameter and high slenderness ratio 

of mini-piles, the load is resisted largely by shaft resistance. The lengths of the rock sockets are 

normally designed to match the pile capacity as limited by the permissible stress of steel bars. A minipile 

usually has four 50 mm diameter high yield steel bars and has a load-carrying capacity of about 

1375 kN. Where mini-piles are installed in soil, the working load is usually less than 700 kN but can be 

in excess of 1 000 kN if post grouting is undertaken using tube-a-manchette. 

Pile cap may be designed to resist horizontal loads. Alternatively, mini-piles can be installed at an 

inclination to resist the horizontal loads. 

c) Large Diameter Bored Piles 

Large-diameter bored piles are used in Malaysia to support heavy column loads of tall buildings and 

highways structures such as viaducts. Typical sizes of these piles range from 1 m to 3 m, with lengths 

up to about 80 m and working loads up to about 45,000 kN. The working load can be increased by 

socketing the piles into rock or providing a bell-out at pile base. The pile bore is supported by 

temporary steel casings or drilling fluid, such as bentonite slurry. 

9-4 March 2009


d) Hand Dug Caissons 

Hand-dug caissons are not very common in Malaysia. For the past two decades, it has been widely 

used in project with limited working space and for hillside development. Their diameters typically range 

from 1.2 m to 2.5 m, with an allowable load of up to about 25000 kN. The advantages and 

disadvantages of hand-dug caissons are summarised in Table 9.2. 

Hand-dug caisson shafts are excavated using hand tools in stages with depths of up to about 1 m, 

depending on the competence of the ground. Dewatering is facilitated by pumping from sumps on the 

excavation floor or from deep wells. Advance grouting may be carried out to provide support in 

potentially unstable ground. Each stage of excavation is lined with in-situ concrete rings (minimum 75 

mm thick) using tapered steel forms which provide a key to the previously constructed rings. When the 

diameter is large, the rings may be suitably reinforced against stresses arising from eccentricity and 

non-uniformity in hoop compression. Near the bottom of the pile, the shaft may be belled out to 

enhance the load-carrying capacity. 

Examples of situations where the use of caissons should be avoided include: 

• Coastal reclamation sites with high groundwater table, 

• Sites underlain by cavernous marble, 

• Deep foundation works (e.g. In excess of say 50 m), 

• Landfill or chemically-contaminated sites, 

• Sites with a history of deep-seated ground movement, 

• Sites in close proximity to water or sewerage tunnels, 

• Sites in close proximity to shallow foundations, and 

• Sites with loose fill having depths in excess of say 10 m. 

Examples of situations where hand-dug caissons may be considered include: 

• Steeply-sloping sites with hand-dug caissons of less than 25 m in depth in soil, and 

• Sites with difficult access or insufficient working room where it maybe impracticable or unsafe 

to use mechanical plant. 

Table 9.2 Advantages and Disadvantages of Hand-dug Caissons 

Advantages 

a) As (a) to (e) for machine-dug piles. 

b) Base materials can be inspected. 

c) Versatile construction method requiring 

minimal site preparation and access. 

d) Removal of obstructions or boulders is 

relatively easy through the use of 

pneumatic drills or, in some cases, 

explosives. 

e) Generally conducive to simultaneous 

excavation by different gangs of workers. 

f) Not susceptible to programme delay arising 

from machine down time. 

g) Can be constructed to large-diameters. 

Disadvantages 

a) As (a), (c) and (e) for machine-dug piles. 

b) Hazardous working conditions for workers 

and the construction method has a poor 

safety record. 

c) Liable to base heave or piping during 

excavation, particularly where the 

groundwater table is high. 

d) Possible adverse effects of dewatering on 

adjoining land and structures. 

e) Health hazards to workers, as reflected by a 

high incidence rate of pneumoconiosis and 

damage to hearing of caisson workers. 

March 2009 9-5


9.2.3 Pile Foundation Design 


Methods based on engineering principles of varying degrees of sophistication are available as a 

framework for pile design. All design procedures can be broadly divided into four categories: 

(a) 

(b) 

(c) 

(d) 

Empirical 'rules-of-thumb', 

Semi-empirical correlations with in-situ test results, 

Rational methods based on simplified soil mechanics or rock mechanics theories, and 

Advanced analytical (or numerical) techniques. 

A judgment has to be made on the choice of an appropriate design method for a given project. 

In principle, in choosing an appropriate design approach, relevant factors that should be considered 

include: 

(a) 

(b) 

(c) 

The ground conditions, 

Nature of the project, and 

Comparable past experience. 

9.2.3.2 Design Philosophies 

The design of piles should comply with the following requirements throughout their service life: 

• There should be adequate safety against failure of the ground. The required factor of safety 

depends on the importance of the structure, consequence of failure, reliability and adequacy of 

information on ground conditions, sensitivity of the structure, nature of the loading, local 

experience, design methodologies, number of representative preliminary pile loading tests. 

• There should be adequate margin against excessive pile movements, which would impair the 

serviceability of the structure. 

a) Global Factor of Safety Approach 

The conventional global factor of safety approach is based on the use of a lumped factor applied 

notionally to either the ultimate strength or the applied load. This is deemed to cater for all the 

uncertainties inherent in the design. 

9-6 March 2009


The conventional approach of applying a global safety factor provides for variations in loads and 

material strengths from their estimated values, inaccuracies in behavioural predictions, unforeseen 

changes to the structure from that analysed, unrecognised loads and ground conditions, errors in 

design and construction, and acceptable deformations in service. 

b) Limit State Design Approach 

A limit state is usually defined as 'any limiting condition beyond which the structure ceases to fulfil 

its intended function'. Limit state design considers the performance of a structure, or structural 

elements, at each limit state. Typical limit states are strength, serviceability, stability, fatigue, 

durability and fire. Different factors are applied to loads and material strengths to account for their 

different uncertainty. 

c) Recommended Factors Of Safety 

The following considerations should be taken into account in the selection of the appropriate 

factors of safety: 

(i) 

(ii) 

(iii) 

(iv) 

(v) 

(vi) 

There should be an adequate safety factor against failure of structural members in 

accordance with appropriate structural codes. 

There must be an adequate global safety factor on ultimate bearing capacity of the ground. 

Terzaghi et al (1991) proposed the minimum acceptable factor of safety to be between 2 

and 3 for compression loading. The factor of safety should be selected with regard to 

importance of structure, consequence of failure, the nature and variability of the ground, 

reliability of the calculation method and design parameters, extent of previous experience and 

number of loading tests on preliminary piles. The factors as summarised in Table 9.3 for 

piles in soils should be applied to the sum of the shaft and end-bearing resistance (HONG 

KONG GEO 2001). 

The assessment of working load should additionally be checked for minimum 'mobilisation' 

factors f s and f b on the shaft resistance and end-bearing resistance respectively as given in 

Table 9.5. 

Settlement considerations, particularly for sensitive structures, may govern the allowable 

loads on piles and the global safety factor and/or 'mobilisation' factors may need to be 

higher than those given in (ii) & (iii) above. 

Where significant cyclic, vibratory or impact loads are envisaged or the properties of the 

ground are expected to deteriorate significantly with time, the minimum global factor of 

safety to be adopted may need to be higher than those in (ii), (iii) and (iv) above. 

Where piles are designed to provide resistance to uplift force, a factor of safety should be 

applied to the estimated ultimate pile uplift resistance and should not be less than the values 

given in Table 9.4. 

March 2009 9-7


Table 9.3 Minimum Global Factors of Safety for Piles in Soil and Rock 

Notes: 

Mobilization Factor for Shaft Mobilization Factor for Endbearing 

Resistance, f b 

Material 

Resistance, f s 

Granular Soils 

1.5 

3 – 5 

Clays 

1.2 

3 – 5 

1. Mobilization factors for end-bearing resistance depend very much on construction. 

Recommended minimum factors assume good workmanship without presences of 

debris giving rise to a ‘soft’ toe and are based on available local instrumented loading 

tests on friction piles in granitic saprolites. Mobilization factors for end-bearing 

resistance. The higher the ratio, the lower is the mobilization factor. 

2. Noting that the movements required to mobilize the ultimate end-bearing resistance 

are about 2% to 5% of the pile diameter for driven piles and about 10% to 20% of 

the pile diameter for bored piles, lower mobilization factor may be used for driven 

piles. 

3. In stiff clays, it is common to limit the peak average shaft resistance to 100 kPa and 

the mobilized base pressure at working load to a nominal value of 550 to 600 kPa for 

settlement considerations, unless higher values can be justified by loading tests. 

4. Where the designer judges that significant mobilization of end-bearing resistance 

cannot be relied on at working load due to possible effects of construction, a design 

approach which is sometimes advocated (e.g. Toh et al, 1989, Brooms & Chang, 1990) 

is to ignore the end-bearing resistance altogether in determining the design working 

load with a suitable mobilization factor on shaft resistance alone (e.g. 1.5). .Endbearing 

resistance is treated as an added safety margin against ultimate failure and 

considered in checking for the factor of safety against ultimate failure. 

5. Lower mobilization factor for end-bearing resistance may be adopted for end-bearing 

piles provided that it can be justified by settlement analyses that the design limiting 

settlement can be satisfied. 

9.2.3.4 Pile Capacity 

a) Design of Geotechnical Capacity in soil 

Pile capacity can be divided into 2 main components, namely; 

• Shaft resistance; Qs 

• End bearing resistance; Qb 

The ultimate capacity of the pile is the sum of both the shaft resistance and the end bearing resistance; 

Qult = Q s + Q b (9.6) 

As for allowable pile capacity; 

Where, 

Q allow = Q s /F s + Q b /Fb (9.7) 

F s = safety factor for shaft resistance. The common F s adopted in design is 2.0 

F b = safety factor for end bearing. The common F b ranges from 2.0 to 3.0 subjected to 

availability and sufficiency of soil parameters. Higher safety factor shall be used when 

limited soil information is made available. As for bored pile, normally Q b is ignored. 

9-8 March 2009


The design of pile geotechnical capacity commonly used can be divided into two major categories 

namely: 

i) Semi-empirical Method 

ii) Simplified Soil Mechanics Method 

i) Semi-Empirical Method 

Piles are constructed in tropical soils that generally have complex soil characteristics. The current 

theoretically based formulae do not consider the effects of soil disturbance, stress relief and partial 

reestablishment of ground stresses that occur during the installation of piles; therefore, the 

sophistication involved in using such formulae may not be necessary. 

Semi-empirical correlations have been extensively developed relating both shaft resistance and base 

resistance of piles to N-values from Standard Penetration Tests (SPT ’N’ values). In the correlations 

established, the SPT ’N’ values generally refer to uncorrected values before pile installation. 

The commonly used correlations for bored piles are as follows: 

f s = K s x SPT ’N’ (in kPa) (9.8) 

f b = K b x SPT ’N’ (in kPa) (9.9) 

Where: 

K s = Ultimate shaft resistance factor 

K b = Ultimate base resistance factor 

SPT’N’ = Standard Penetration Tests blow counts (blows/300mm) 

Toh et al. (1989) reported that the average K s obtained varies from 5 at SPT ’N’ 20 to as low as 1.5 at 

SPT ’N’=220. Chang & Broms (1991) suggests that K s of 2 for bored piles in residual soils of Singapore 

with SPT ’N’


Where : 

α = adhesion factor 

s u = undrained shear strength (kPa) 

Whitaker & Cooke (1911) reports that the α value lies in the range of 0.3 to 0.1 for stiff 

overconsolidated clays, while Tomlinson (1994) and Reese & O’Neill (1988) report α values in the range 

of 0.4 to 0.9. The α values for residual soils of Malaysia are also within this range. Where soft clay is 

encountered, a preliminary value of 0.8 to 1.0 is usually adopted together with the corrected 

undrained shear strength from the vane shear test. This method is useful if the bored piles are to be 

constructed on soft clay near river or at coastal area. 

The value of ultimate shaft resistance can also be estimated from the following expression: 

f s = Kse x σv ’ x tan φ’ (9.11) 

Where : 

Kse = Effective Stress Shaft Resistance Factor = [can be assumed as Ko] 

σv ’ = Vertical Effective Stress (kPa) 

φ’ = Effective Angle of Friction (degree) of fined grained soils. 

However, this method is not popular in Malaysia and limited case histories of back-analysed K se values 

are available for practical usage of the design engineer. 

Although the theoretical ultimate base resistance for pile in fine grained soil can be related to 

undrained shear strength as follows; 

f b = N c x s u (9.12) 

Where: 

N c = bearing capacity factor 

Note: it is not recommended to include base resistance in the calculation of the bored pile geotechnical 

capacity due to difficulty and uncertainty in base cleaning. 

Coarse Grained Soils 

The ultimate shaft resistance (f s ) of piles in coarse grained soils can be expressed in terms of effective 

stresses as follows: 

f su = β x σv’ (9.13) 

Where: 

β = shaft resistance factor for coarse grained soils. 

The β values can be obtained from back-analyses of pile load tests. The typical β values of piles in 

loose sand and dense sand are 0.15 to 0.3 and 0.25 to 0.1 respectively based on Davies & Chan 

(1981). 

9-10 March 2009


c) Negative Skin Friction 

Piles installed through compressible materials (e.g. fill or marine clay) can experience negative skin 

friction. This occurs on the part of the shaft along which the downward movement of the 

surrounding soil exceeds the settlement of the pile. Negative skin friction could result from 

consolidation of a soft deposit caused by dewatering or the placement of fill. The dissipation of excess 

pore water pressure arising from pile driving in soft clay can also result in consolidation of the clay. 

The magnitude of negative skin friction that can be transferred to a pile depends on (Bjerrum, 

1973): 

(a) 

(b) 

(c) 

(d) 

Pile material, 

Method of pile construction, 

Nature of soil, and 

Amount and rate of relative movement between the soil and the pile 

In determining the amount of negative skin friction, it would be necessary to estimate the position of 

the neutral plane, i.e. the level where the settlement of the pile equals the settlement of the 

surrounding ground. For end-bearing piles, the neutral plane will be located close to the base of the 

compressible stratum. 

Calculation of Negative Skin Friction 

Design of negative skin friction should include checks on the structural and geotechnical capacity of 

the pile, as well as the downward movement of the pile due to the negative skin friction dragging 

the pile shaft (CGS, 1992; Fellenius, 1998). A pile will settle excessively when geotechnical failure 

occurs. As the relative displacement between the soil and the pile shaft is reversed, the effect of 

negative skin friction on pile shaft would be eliminated. Therefore, the geotechnical capacity of the 

pile could be based on the shaft resistance developed along the entire length of pile. The drag load 

need not be deducted from the assessed geotechnical capacity when deciding the allowable load 

carrying capacity of the pile. On the other hand, the structural capacity of the pile should be sufficient 

to sustain the maximum applied load and the drag load. The drag load should be computed for a 

depth starting from the ground surface to the neutral plane. 

The estimation of downward movement of the pile (i.e. downdrag) requires the prediction of the 

neutral plane and the soil settlement profile. At the neutral plane, the pile and the ground settle by 

the same amount. The neutral plane is also where the sustained load on the pile head plus the 

dragload is in equilibrium with the positive shaft resistance plus the toe resistance of the pile. The 

total pile settlement can therefore be computed by summing the ground settlement at the neutral 

plane and the compression of the pile above the neutral plane (Figure 9.2). For piles founded on 

a relatively rigid base (e.g. on rock) where pile settlement is limited, the problem of negative skin 

friction is more of the concern on the structural capacity of the pile. 

This design approach is also recommended in the Code of Practice for Foundations (BD, 2004a) for 

estimating the effect of negative skin friction. 

For friction piles, various methods of estimating the position of the neutral plane, by determining the 

point of intersection of pile axial displacement and the settlement profile of the surrounding soil, 

have been suggested by a number of authors (e.g. Fellenius, 1984). However, the axial 

displacement at the pile base is generally difficult to predict without pile loading tests in which the 

base and shaft responses have been measured separately. The neutral plane may be taken to be 

the pile base for an end-bearing pile that has been installed through a thick layer of soft clay down 

March 2009 9-11


to rock or to a stratum with high bearing capacity. The method includes the effect of soil- structure 

interaction in estimating the neutral plane and drag load on a pile shaft. Alternatively, the neutral 

plane can be conservatively taken as at the base of the lowest compressible layer (BD, 2004a). 

The mobilised negative skin friction, being dependent on the horizontal stresses in the ground, will be 

affected by the type of pile. For steel H-piles, it is important to check the potential negative skin 

friction with respect to both the total surface area and the circumscribed area relative to the available 

resistance (Broms, 1979). 

The effective stress or β method may be used to estimate the magnitude of negative skin friction on 

single piles (Bjerrum et al, 1919; Burland & Starke, 1994). 

In general, it is only necessary to take into account negative skin friction in combination with dead 

loads and sustained live load, without consideration of transient live load or superimposed load. 

Transient live loads will usually be carried by positive shaft resistance, since a very small 

displacement is enough to change the direction of the shaft resistance from negative to positive, 

and the elastic compression of the piles alone is normally sufficient. In the event where the 

transient live loads are larger than twice the negative skin friction, the critical load condition will be 

given by (dead load + sustained live load + transient live load). The above recommendations are 

based on consideration of the mechanics of load transfer down a pile (Broms, 1979) and the 

research findings (Bjerrum et al, 1919; Fellenius, 1972) that very small relative movement will be 

required to build up and relieve negative skin friction, and elastic compression of piles associated 

with the transient live load will usually be sufficient to relieve the negative skin friction. Caution 

needs to be exercised however in the case of short stubby piles founded on rock where the elastic 

compression may be insufficient to fully relieve the negative skin friction. In general, the customary 

local assumption of designing for the load combination of (dead load + full live load + negative skin 

friction) is on the conservative side. 

Poulos (1990b) demonstrated how pile settlement can be determined using elastic theory with 

due allowance for yielding condition at the pile/soil interface. If the ground settlement profile is 

known with reasonable certainty, due allowance may be made for the portion of the pile shaft over 

which the relative movement is insufficient to fully mobilise the negative skin friction (i.e. movement 

less than 0.5% to 1% of pile diameter). 

9-12 March 2009


Notes: 

(1) The negative skin friction, f n , in granular soils and cohesive soils is determined as 

for positive shaft resistance, the effective stress approach can be used to 

estimate the negative skin friction as follows: 

f n = ßσ v ’ 

where 

f n = negative skin friction 

σ v ’ = vertical effective stress 

ß = empirical factor obtained from full-scale loading tests or based on the 

soil mechanics principle. 

(2) Ultimate load-carrying capacity of pile will be mobilized when pile settles more than 

the surrounding soil. In such case, the geotechnical capacity of the 

pile can be 

calculated based on 

the entire length of pile. 

Figure 9.2 Estimation of Negative Skin Friction by Effectivee Stress Method 

9.2.4 

9.2.4.1 

Pile Loading Tests 

General 

Given the many uncertainties in the design and construction of piles, it is difficult 

to accurately 

predict the performance of a pile. Loading tests can be carried out on preliminary 

piles to confirm 

the pile design capacity or on working piles as a proof loading tests. Although pile loading tests 

add to the cost of foundation, the 

saving can be significant in the event that improvement of to the 

foundation design can 

be materialised. 

March 2009 

9-13


There are two main types of pile loading tests, namely static and dynamic loading tests. Static 

loading tests are generally preferred because they have been traditionally used and also because 

they are perceived to replicate the long-term sustained load conditions. Dynamic loading tests are 

usually carried out as a supplement to static loading tests and are generally less costly when 

compared with static loading tests. The failure mechanism in a dynamic loading test may be 

different from that in a static loading test. 

The Statnamic loading test is a quasi-static loading test with limited local experience. In this test, a 

pressure chamber and a reaction mass is placed on top of the pile. Solid fuel is injected and burned 

in the chamber to generate an upward force on the reaction mass. An equal and opposite force 

pushes the pile downward. The pile load increases to a maximum and is then reduced when 

exhausted gases are vented from the pressure chamber. 

Pile displacement and induced force are automatically recorded by laser sensors and a load cell. The 

load duration for a Statnamic loading test is relatively long when compared with other high energy 

dynamic loading tests. While the additional soil dynamic resistance is usually minimal and a 

conventional static load-settlement curve can be produced, allowance will be required in some soil 

types such as soft clays. 

9.2.4.2 Timing of Pile Tests 

For cast-in-place piles, the timing of a loading test is dictated by the strength of the concrete or 

grout in the pile. Weltman (1980b) recommended that at the time of testing, the concrete or grout 

should be a minimum of seven (7) days old and have strength of at least twice the maximum applied 

stress. 

With driven piles, there may be a build-up of pore water pressure after driving. Lam et al (1994) 

reported that for piles driven into weathered meta-siltstone the excess pore water pressure built up 

during driving took only one and a half days to dissipate completely. 

Results of dynamic loading tests reported by Ng (1989) for driven piles in loose granitic 

saprolites (with SPT N values less than 30) indicated that the measured capacities increased by 

15% to 25% in the 24 hours after installation. The apparent 'set up' may have resulted from 

dissipation of positive excess pore water pressure generated during pile driving. 

As a general guideline, a driven pile should be tested at least three days after driving if it is driven 

into a granular material and at least four weeks after driving into a clayey soil, unless sufficient local 

experience or results of instrumentation indicate that a shorter period would be adequate for 

dissipation of excess pore pressure. 

9.2.4.3 Static Pile Loading Tests 

a) Reaction Arrangement 

To ensure stability of the test assembly setup, careful consideration should be given to the provision 

of a suitable reaction system. The geometry of the arrangement should also aim to minimise 

interaction between the test pile, reaction system and reference beam supports. It is advisable to 

have, say, a minimum of 20% margin on the capacity of the reaction against maximum test load. 

9-14 March 2009


i) Compression tests 

Kentledge is commonly used in Malaysia as the reaction system (Figure 9.3). This involves the use 

of dead weights (comprises of concrete blocks) supported by a deck of steel beams sitting 

on crib pads. The area of the crib should be sufficient to avoid bearing failure or excessive 

settlement of the ground. It is recommended that the crib pads are placed at least 1.3 m from 

the edge of the test pile to minimise interaction effects. If the separation distance is less than 

1.3 m, the surcharge effect from the kentledge should be determined and allowed for in the 

interpretation of the loading test results. 

Sometimes tension piles are used to provide reaction for the applied load (Figure 9.4) and 

should be located as far as practicable from the test pile to minimise interaction effects. A 

minimum centre-to-centre spacing of 2 m or three pile diameters, whichever is greater, between the 

test pile and tension piles is recommended. If the centre spacing between piles is less than three 

pile diameters, there may be significant pile interaction and the observed settlement of the test 

pile will be less than what should have been. If a spacing of less than three pile diameters is adopted, 

uplift of the tension piles should be monitored and corrections should be made for the settlement of 

the test pile based on recognised methods considering pile interaction. A minimum of three reactions 

piles should be used to prevent instability of the set up during pile loading tests. Alternatively some 

from of lateral support should be provided. 

Figure 9.3 Typical Arrangement of a Compression Test using Kentledge 

March 2009 9-15


Figure 9.4 Typical Arrangement of a Compression Test using Tension Piles 

To reduce interaction between the ground anchors and the test pile, the fixed lengths of the anchors 

should be positioned a distance away from the centre of the test pile of at least three pile of diameters 

or 2 m, whichever is greater. Ground anchors may be used instead of tension piles to provide load 

reaction. The main shortcomings with ground anchors are the tendon flexibility and their vulnerability 

to lateral instability. 

The provision of a minimum of four ground anchors is preferred for safety considerations. 

Installation and testing of each ground anchor should be in accordance with the recommendations 

as given in GCO (1989) for temporary anchors. The anchor load should be locked off at 110% 

design working load. The movements of the anchor should be monitored during the loading tests to 

give prior warning of any imminent abrupt failure. 

The use of ground anchors will generally be most suitable in testing a raking pile because the 

horizontal component of the jacking may not be satisfactorily restrained in other reaction systems. 

They should be inclined along the same direction as the raking pile. 

9-16 March 2009


Traditionally, a static loading test is carried out by jacking a pile against a kentledge or a reaction 

frame supported by tension piles or ground anchors. In recent years, Osterberg load cell (O-cell) has 

been widely adopted for static loading tests for large-diameter cast-in- place concrete piles. It can 

also be used in driven steel piles. 

An O-cell is commonly installed at or near the bottom of the pile. Reaction to the upward force 

exerted by the O-cell is provided by the shaft resistance. For such testing arrangement, the shaft 

resistance mobilised in the pile will be in upward direction. A smaller kentledge may be assembled 

in case the shaft resistance alone is not adequate to resist the applied load. The maximum test 

load is governed by either the available shaft resistance, the bearing stress at the base or the 

capacity of the O-cell itself. 

ii) 

Uplift loading tests 

A typical arrangement for uplift loading tests is shown in Figure 9.4. The arrangement involving 

jacking at the centre is preferred because an even load can be applied 

Reaction piles should be placed at least three test pile diameters, or a minimum of 2 m, from the 

centre of the test pile. Where the spacing is less than this, corrections for possible pile interaction 

should be made. Alternatively, an O-cell installed at the base of pile can also be used in an uplift 

test. 

iii) Lateral loading tests 

In a lateral loading test, two piles or pile groups may be jacked against each other (Figure 

9.5(a)). It is recommended that the centre spacing of the piles should preferably be a minimum 

of ten pile diameters (CGS, 1992). 

Alternative reaction systems including a 'deadman' or weighted platform are also shown in 

Figure 9.5(b) and (c). 

9.2.5 Equipment 

9.2.5.1 Measurement of Load 

A typical load application and measurement system consists of hydraulic jacks, load measuring 

device, spherical seating and load bearing plates (Figure 9.3). 

The jacks used for the test should preferably be large-diameter low-pressure jacks with a travel of 

at least 15% of the pile diameter (or more if mini-piles are tested). A single jack is preferred where 

practicable. If more than one jack is used, then the pressure should be applied using a motorised 

pumping unit instead of a hand pump. Pressure gauges should be fitted to permit a check on the 

load. The complete jacking system including the hydraulic cylinder, valves, pump and pressure 

gauges should be calibrated as a single unit. 

It is strongly recommended that an independent load-measuring device in the form of a load cell, 

load column or pressure cell is used in a loading test. The device should be calibrated before 

each series of tests to an accuracy of not less than 2% of the maximum applied load. 

It is good practice to use a spherical seating in between the load measuring device and bearing plates 

in a compression loading test in order to minimise angular misalignment in the system and ensure 

that the load is applied coaxially to the test pile. Spherical seating is however only suitable for 

correcting relatively small angular misalignment of not more than about 3°. 

March 2009 9-17


A load bearing plate should be firmly bedded onto the top of the pile (or the pile cap) orthogonal to 

the direction of applied load so as to spread the load evenly onto the pile. An O-cell consists of two 

steel plates between which there is an expandable pressurised chamber. Hydraulic fluid is injected 

to expand the chamber, which pushes the pile segment upward. At the same time, the bearing base 

(or lower pile segment if the O-cell is installed in middle of the pile) is loaded in the downward 

direction. Pressure gauges are attached to fluid feed lines to check the applied load and it is 

necessary to calibrate the O-cell. Correction may be needed to allow for the level difference 

between the pressure gauges, which is located at the ground surface and the load cell, which is 

usually installed at the base of the piles. 

9-18 March 2009


Figure 9.5 Typical Arrangement of a Lateral Loading Test 

9.2.5.2 

Measurement of Pile Head Movement 

Devices used for measuring pile head settlement in a loading test include dial gauges (graduated to 

0.01 mm), linear variable differential transducers (LVDT) and optical levelling systems. A system 

consisting of a wire, mirror and scale is also used in lateral loading tests. 

March 2009 

9-19


In a compression or tension test, measurements should be taken by four dial gauges evenly spaced 

along the perimeter of the pile to determine whether the pile head tilts significantly. The measuring 

points of the gauges should sit on the pile head or on brackets mounted on the side of the pile with 

a glass slide or machined steel plate acting as a datum for the stems. Care should be taken to 

ensure that the plates are perpendicular to the pile axis and that the dial gauge stems are in line with 

the axis. 

In a lateral loading test, dial gauges should be placed on the back of the pile with the stems in line 

with the load for measuring pile deflection (Figure 9.5). A separate system involving the use of a 

wire, mirror and scale may be used as a check on the dial gauges. The wire should be held under 

constant tension and supported from points at a distance not less than five pile diameters from the 

test pile and any part of the reaction system. 

Rotational and transverse movement of the pile should also be measured. 

LVDT can be used in place of dial gauges and readings can be taken remotely. However, they 

are susceptible to dirt and should be properly protected in a test. 

The reference beams to which the dial gauges or LVDT are attached should be rigid and stable. A 

light lattice girder with high stiffness in the vertical direction is recommended. This is better than 

heavy steel sections of lower rigidity. To minimise disturbance to the reference beams, the 

supports should be firmly embedded in the ground away from the influence of the loading 

system (say 2 m from piles or 1 m from kentledge support). It is recommended that the beam is 

clamped on one side of the support and free to slide on the other. Such an arrangement allows 

longitudinal movement of the beam caused by changes in temperature. The test assembly should be 

shaded from direct sunlight. 

In an axial loading test, levels of the test pile and reference beam supports should be monitored by 

an optical levelling system throughout the test to check for gross errors in the measurements. The 

optical levelling should be carried out at the maximum test load of each loading cycle and when the 

pile is unloaded at the end of each cycle. The use of precision levelling equipment with an accuracy 

of at least 1 mm is preferred. The datum for the optical levelling system should be stable and 

positioned sufficiently far away from the influence zone of the test. 

In loading tests using O-cell, rod extensometers are connected to the top and bottom plates of the 

O-cell (Figure 9.6). They are extended to the ground surface such that the movement of the 

plates can be measured by dial gauges or displacement transducers independently. 

9-20 March 2009


Figure 9.5 Typical Instrumentation Scheme for a Vertical Pile Loading Test 

9.2.5.3 

Test Procedures 

a) General 

Two types of loading test procedures are commonly used, namely 

maintained-load (ML) and 

constann t-rate-of-penetration (CRP) tests. The ML method is applicable t o compression, 

tension and lateral loading tests, 

whereas the CRP method is used mainly in compression loading 

tests. 

The design working load (W L 

) of the pile should be pre-determined 

where W L 

is defined as the 

allowable 

load for a pile before allowing for 

factors such as negative skin friction, group effects 

and redundancy. 

March 2009 

9-21


b) Maintained-load tests 

In a maintained-load test, the load is applied in increments, each being held until the rate of 

movement has reduced to an acceptably low value before the next load increment is applied. It is 

usual practice to include a number of loading and unloading cycles in a loading test. Such cycles can 

be particularly useful in assessing the onset of plastic movements by observing development of the 

residual (or plastic) movement with increase in load. 

Details of the common loading procedures used in Hong Kong GEO which can be used as a guide are 

summarised in Table 9.4. 

When testing a preliminary pile, the pile should, where practicable, be loaded to failure or at 

least to sufficient movement (say, a minimum of 5% of pile diameter). If the pile is loaded beyond 2 

W L 

, a greater number of small load increments, of say 0.15 to 0.2 W L 

as appropriate, may be used 

in order that the load-settlement behaviour can be better defined before pile failure. However, 

the test load should not exceed the structural capacity of the pile. 

In principle, the same loading procedures suggested for compression tests may be used for 

lateral and uplift loading tests. 

9-22 March 2009


Table 9.4 Loading Procedures and Acceptance Criteria for Pile Loading Tests in Hong Kong 

General 

Specification for 

Civil Engineering 

Works (HKG, 1992) 

Cycle 1-25% Q max 



1. δ Q 1.8W L ). 

3. Load increments/ 

decrements not to be 

applied until rate of 

settlement or rebound of 

pile is less than 0.1 mm in 

20 minutes. 

4. Full load at each cycle to 

be maintained for at least 

24 hours after rate of 

settlement has reduced to 

less than 0.1 mm per hour. 

Code of Practice 

for Foundations 

(BD, 2004a) 

Loading schedule 

for piles with a 

diameter or at least 

lateral dimension 

not exceeding 750 

mm: 

Cycle 1 – 100% W L 

Cycle 2 – 200% W L 

(=Q max ) 

1. δ max < Q L 

D 

4 

A E 

(mm) 

2. The greater of: 

δ max < D 

4 or 

 

0.25δ max (in mm) 

Legend : δ Q 

= pile head settlement at failure or maximum test load 

δ 90%Q 

= pile head settlement at 90% of failure or maximum test load 

δ max 

= maximum pile head settlement 

δ = pile head settlement 

1. Load increments/ 

decrements to be in 50% 

of the design working load; 

pile to be unloaded at the 

end of each cycle. 

2. Piles are to e tested to 

twice design working load. 

3. Increments of load not to 

be applied until rate of 

settlement or recovery of 

pile is less than 0.05 mm in 

10 minutes. 

4. Full load at cycle 2 should 

be maintained for at least 

72 hours. 

5. The residual settlement, 

δ res , should be taken when 

the rate of recovery of the 

pile after removal of test 

load is less than 0.1 mm in 

15 minutes. 

March 2009 9-23


δ res 

= residual (or permanent) pile head settlement upon unloading from maximum 

load 

Q max 

= maximum test load 

W L 

= design working load of pile 

L = pile length 

Ap = cross-sectional area of pile 

Ep = Young's modulus of pile 

D = least lateral dimension of pile section (mm) 

9.2.5.4 Instrumentation 

a) General 

Information on the load transfer mechanism can be derived from a loading test if the pile is 

instrumented. To ensure that appropriate and reliable results can be obtained, the pile 

instrumentation system should be compatible with the objectives of the test. Important 

aspects including selection, disposition and methods of installation should be carefully considered. 

It is essential that sufficient redundancy is built in to allow for possible damage and 

malfunctioning of instruments. Where possible, isolated measurements i.e., survey leveling method 

should be made using more than one type of equipment to permit cross-checking of results. An 

understanding of the ground profile, proposed construction technique and a preliminary 

assessment of the probable behaviour of the pile will be helpful in designing the disposition of the 

instruments. Limitations and resolutions of the instruments should be understood. 

b) Axial loading tests 

Information that can be established from an instrumented axial loading test includes the 

distribution of load and movement, development of shaft resistance and end-bearing 

resistance with displacement. A typical instrumentation layout is given in Figure 9.6. 

Strain gauges (electrical resistance and vibrating wire types) can be used to measure local 

strains, which can be converted to stresses or loads. Vibrating wire strain gauges are generally 

preferred, particularly for long-term monitoring, as the readings will not be affected by changes in 

voltage over the length of cable used, earth leakage, corrosion to connection and temperature 

variation. In case measurements need to be taken rapidly, e.g. in simulation dynamic response of 

piles, electrical resistance type strain gauges are more suitable. 

A variant form of vibrating wire strain gauges is the 'sister bar' or 'rebar strain meter'. This is 

commonly used in cast-in-place concrete piles. It consists of a vibrating strain gauge assembled 

inside a high strength steel housing that joins two reinforcement bars at both ends by welding or 

couplers. The sister bar can replace a section of the steel in the reinforcement cage or be placed 

alongside it. Such an arrangement minimises the chance that a strain gauge is damaged during 

placing of concrete. The electrical wirings should be properly tied to the reinforcement cage at 

regular intervals. 

To measure axial loads, the strain gauge stems are orientated in line with the direction of the load 

(i.e. vertical gauges). One set of gauges should be placed near the top of the pile, and preferably 

in a position where the pile shaft is not subject to external shaft resistance, to facilitate calculation 

of the modulus of the composite section. Gauges should also be placed close to the base of the 

9-24 March 2009


pile (practically 0.5 m) with others positioned near stratum boundaries and at intermediate levels. A 

minimum of two and preferably four gauges should be provided at each level where practicable. 

c) Lateral loading tests 

The common types of internal instrumentation used in a lateral loading test are inclinometers, strain 

gauges and electro-levels. 

The deflected shape of a pile subject to lateral loading can be monitored using an 

inclinometer. The system consists of an access tube and a torpedo sensor. For cast-in-place 

piles, the tube is installed in the pile prior to concreting. For displacement piles such as H- piles, 

a slot can be reserved in the pile by welding on a steel channel or angle section prior to pile 

driving. The tube is grouted into the slot after driving. During the test, a torpedo is used to 

measure the slope, typically in 0.5 m gauge lengths, which can be converted to deflections. 

Care needs to be exercised in minimising any asymmetrical arrangement of the pile section or 

excessive bending of the pile during welding of the inclinometer protective tubing. In 

extreme cases, the pile may become more prone to being driven off vertical because of these 

factors. 

Strain gauges with their stems orientated in line with the pile axis can be used for measuring 

direct stresses and hence bending stresses in the pile. They can also be oriented horizontally to 

measure lateral stresses supplemented by earth pressure cells. 

Electro-levels measure changes in slope based on the inclination of an electrolytic fluid that 

can move freely relative to three electrodes inside a sealed glass tube (Price & Wardle, 1983; 

Chan & Weeks, 1995). The changes in slope can be converted to deflections by multiplying the 

tangent of the change in inclination by the gauge length. The devices are mounted in an 

inclinometer tube cast into the pile and can be replaced if they malfunction after installation. 

Earth pressure cells can also be used to measure the changes in normal stresses acting on the pile 

during loading. It is important that these pressure cells are properly calibrated for cell action 

factors, etc. to ensure sensible results are being obtained. 

9.2.5.5 Interpretation of Test Results 

a) General 

A considerable amount of information can be derived from a pile loading test, particularly with an 

instrumented pile. In the interpretation of test results for design, it will be necessary to consider 

any alterations to the site conditions, such as fill placement, excavation or dewatering, which can 

significantly affect the insitu stress level, and hence the pile capacity, after the loading test. 

b) Evaluation of failure load 

Typical load-settlement curves, together with some possible modes of failure, are shown in 

Figure 9.7. Problems such as presence of a soft clay layer, defects in the pile shaft and poor 

construction techniques may be deduced from the curves where a pile has been tested to 

failure. 

March 2009 9-25


It is difficult to definee the failure load of a pile 

when it has 

not been loaded to failure. In the case 

where ultimate failure has not been reached in a loading 

test, a limiting load may be defined 

which corresponds to 

a limiting settlement or rate of settlement. A commonly-used definition of 

failure load is taken 

to be that at which settlement continues to 

increase without further 

increase in load; alternatively, it is customarily taken as the load causing a settlement of 10% 

of pile diameter (BSI, 1981). However, it should be noted that elastic shortening 

of very long 

pile can already exceed 10% of the pile diameter. O'Neill & Reese (1999) suggested using the 

load thatt gives a pile 

head settlement of 5% of the diameter of bored piles as the ultimate end- 

bearing capacity, if failure does not occur. It 

is also recommended to take the failure load to be 

the load that gives a pile head settlement of 

4.5% of the pile diameter plus 75% 

of the elastic 

shortening of pile. In 

practice, the 

failure or ultimate load represents no 

more than a benchmark 

such that the safe design working load can be determinedd by applying 

a suitable factor of safety. 

Figure 9.6 Typical Load Settlement Curves for Pile Loading Testss (Tomlinson, 1994) 

c) Acceptance criteria 

From the load-settlement curve, 

a check of pile acceptability in terms of compliance with 

specified criteria can 

be made. It is recommended that the acceptance criteria given in Code of 

Practice for Foundation (BD 2004) to be adopted. 

9-26 

March 2009


The acceptance criteria specified in the Code of Practice for Foundations (BD, 2004a) are generally 

adopted for engineering practice in Malaysia. 

Non-compliance with the criterion on acceptance criteria does not necessarily imply nonacceptance 

of the pile. Where this criterion is not met, it is prudent to examine the pile behaviour 

more closely to find out the reasons of non-compliance. 

In principle, a designer should concentrate on the limiting deflection at working load as well as 

the factor of safety against failure or sudden gross movements. The limiting settlement of a 

test pile at working load should be determined on an individual basis taking into account the 

sensitivity of the structure, the elastic compression component, effects of pile group interaction 

under working condition, and expected behaviour of piles as observed in similar precedents. 

In analysing the settlement behaviour of the pile under a pile loading test, it is worth noting that 

the applied load will be carried in part or entirely by the shaft resistance, although the shaft 

resistance may be ignored in the pile design. Consequently, the elastic compression component of 

pile could be smaller than that estimated based on the entire length of the pile, particularly for 

long friction pile. Fraser & Ng (1990) suggested that upon removal of the maximum test load, 

the recovery of the pile head settlement may be restricted by the 'locked in' stress as a result of 

reversal of shaft resistance upon removal of the test load. 

9.2.6 Dynamic Loading Tests 


Various techniques for dynamic loading tests are now available. These tests are relatively 

cheap and quick to carry out compared with static loading tests. Information that can be obtained 

from a dynamic loading test includes: 

(a) static load capacity of the pile, 

(b) energy delivered by the pile driving hammer to the pile, 

(c) maximum driving compressive stresses (tensile stress should be omitted), 

(d) location and extent of structural damage. 

9.2.6.2 Test Methods 

The dynamic loading test is generally carried out by driving a prefabricated pile or by applying 

impact loading on a cast-in-place pile by a drop hammer. A standard procedure for carrying out a 

dynamic loading test is given in ASTM (1995b). 

The equipment required for carrying out a dynamic pile loading test includes a driving hammer, 

strain transducers and accelerometers, together with appropriate data recording, processing 

and measuring equipment. 

The hammer should have a capacity large enough to cause sufficient pile movement such that 

the resistance of the pile can be fully mobilised. A guide tube assembly to ensure that the force 

is applied axially on the pile should be used. 

The strain transducers contain resistance foil gauges in a full bridge arrangement. The 

accelerometers consist of a quartz crystal which produces a voltage linearly proportional to the 

acceleration. A pair of strain transducers and accelerometers are fixed to opposite sides of the 

pile, either by drilling and bolting directly to the pile or by welding mounting blocks, and 

March 2009 9-27


positioned at least two diameters or twice the length of the longest side of the pile section below 

the pile head to ensure a reasonably uniform stress field at the measuring elevation. 

In the test, the strain and acceleration measured at the pile head for each blow are recorded. 

The signals from the instruments are transmitted to a data recording, filtering and displaying 

device to determine the variation of force and velocity with time. 

9.2.6.3 Methods of Interpretation 

a) General 

Two general types of analysis based on wave propagation theory, namely direct and indirect 

methods, are available. Direct methods of analysis apply to measurements obtained directly from 

a (single) blow, whilst indirect methods of analysis are based on signal matching carried out on 

results obtained from one or several blows. 

Examples of direct methods of analysis include CASE, IMPEDANCE and TNO method, and indirect 

methods include CAPWAP, TNOWAVE and SIMBAT, CASE and CAPWAP analyses are used 

mainly for displacement piles, although in principle they can also be applied to cast-in-place 

piles. SIMBAT has been developed primarily for cast-in- place piles, but it is equally applicable to 

displacement piles. 

In a typical analysis of dynamic loading test, the penetration resistance is assumed to be 

comprised of two parts, namely a static component, Rs, and a dynamic component, Rd. 

b) CAPWAP method 

CAPWAP (CAse Pile Wave Analysis Program) analysis is the common analysis adopted by the local 

tester in Malaysia. In a CAPWAP analysis, the soil is represented by a series of elasto-plastic 

springs in parallel with a linear dashpot similar to that used in the wave equation analysis 

proposed by Smith (1912). The soil can also be modelled as a continuum when the pile is 

relatively short. CAPWAP measures the acceleration-time data as the input boundary condition. 

The program computes a force versus time curve which is compared with the recorded data. If 

there is a mismatch, the soil model is adjusted. This iterative procedure is repeated until a 

satisfactory match is achieved between the computed and measured force-time diagrams. 

The dynamic component of penetration resistance is given by: 

R d = j s v p R s (9.14) 

Where: 

j s = Smith damping coefficient 

v p = velocity of pile at each segment 

R s = static component of penetration resistance 

Input parameters for the analysis include pile dimensions and properties, soil model parameters 

including the static pile capacity, Smith damping coefficient, js and soil quake (i.e. the amount of 

elastic deformation before yielding starts), and the signals measured in the field. The output will 

be in the form of distribution of static unit shaft resistance against depth and base response, 

together with the static load-settlement relationship up to about 1.5 times the working load. It 

should be noted that the analysis does not model the onset of pile failure correctly and care should 

be exercised when predicting deflections at loads close to the ultimate pile capacity. 

9-28 March 2009


9.2.6.4 Recommendations on the Use of Dynamic Loading Tests 

Traditionally, pile driving formulae are used as a mean to assess pile capacity from a measurement 

of 'set per blow' and are supplemented with static loading tests on selected piles. Although such an 

approach is the standard in local practice for driving piles, driving formulae are considered 

fundamentally incorrect and quantitative agreement between static pile capacities predicted by 

driving formulae and actual values cannot be relied upon. 

Dynamic load testing is preferred for pile capacity predictions. Dynamic load testing can be 

applied to non-homogeneous soils or piles with a varying cross-sectional area. The static loadsettlement 

response of a pile can also be predicted. 

Dynamic pile loading tests can supplement the design of driven piles provided that they have 

been properly calibrated against static loading tests and an adequate site investigation has been 

carried out. It should be noted that such calibration of the analysis model has to be based on 

static loading tests on piles of similar length, cross section and under comparable soil conditions 

and loaded to failure. A static loading test, which is carried out to a proof load, is an inconclusive 

result for assessing the ultimate resistance of the pile. 

The reliability of the prediction of dynamic loading test methods is dependent on the adequacy of 

the wave equation model and the premise that a unique solution exists when the best fit is 

obtained within the limitation of the assumption of an elasto/rigid plastic soil behavior. In 

addition, there are uncertainties with the modelling of effects of residual driving stresses in the 

wave equation formulation. 

9.3 LATERALLY LOADED PILES 

9.3.1 Introduction 

The lateral load capacity of a pile may be limited by the following: 

(a) Shear capacity of the soil; 

(b) Structural (i.e. bending moment and shear) capacity of the pile section itself; and 

(c) Excessive deformation of the pile. 

The failure mechanisms of short piles under lateral loads as compared to those of long piles differ, 

requiring therefore different and appropriate design methods. In order to establish if a pile behaves 

a rigid unit (i.e. short pile) or as a flexible member (i.e. long pile), the stiffness factors as defined in 

Figure 9.8 below will employed. 

March 2009 9-29


Notes: 

1. For constant soil modulus with depth (e.g. stiff overconsolidated clay), pile stiffness factor 

4 

R = E pI p 

(in units of length) where E p I p is the bending stiffness of the pile, D is the 

k h D 

width of the pile, k h is the coefficient of horizontal subgrade reaction. 

2. For soil modulus increases linearly with depth (e.g. normally consolidated clay & granular 

5 

soils), pile stiffness factor, T = E pI p 

where n h is the constant of horizontal subgrade 

n h 

reaction given in table below: 

3. The criteria for behaviour as a short (rigid) pile or as a long (flexible) pile are as follows: 

Pile Type 

Short (rigid) piles 

Long (flexible) piles 

Soil Modulus 

Linearly increasing 

L 2T 

L 4T 

Constant 

L 2R 

L 3.5R 

Figure 9.7 Failure Modes of Vertical Piles under Lateral Loads (Broms, 1914a) 

9-30 March 2009


Consistency 

(MN/m 3 ) 

Table 9.5 Typical Values of Coefficient of Horizontal Subgrade Reaction 

Loose 

(N value 4- 

Medium Dense 

(N value 11- 

n h for dry or 2.2 6.6 17.6 

n h for 1.3 4.4 10.7 

Dense 

(N value 31- 

Notes: 

i. The above n h values are based on Terzaghi (1955) and are valid for stresses up to about half the 

ultimate bearing capacity with allowance made for long-term movements. 

ii. For sands, Elson (1984) suggested that Terzaghi's values should be used as a lower limit and the 

following relationship as the upper limits : 

n h = 

where D r is the relative density of sand in percent. 

iii. Other observed values of n h , which include an allowance for long-term movement, are as follows 

(Tomlinson, 1994) : 

Soft normally consolidated clays: 350 to 700 

Soft organic silts: 150 kN/m 3 

iv. For sands, n h may be related to the drained horizontal Young modulus (E h ') in MPa as follows 

(Yoshida & Yoshinaka, 1972; Parry, 1972) : 

n h = 0.8 h ' to 1.8 h 

' 

z 

(9.16) 

where z is depth below ground surface in metres. 

v. It should be noted that empirical relationships developed for transported soils between N value 

and relative density are not generally valid for weathered rocks. Corestones, for example, can 

give misleading high values that are unrepresentative of the soil mass. 

As the surface soil layer can be subject to disturbance, suitable allowance should be made in the 

design by ignoring as appropriate, the resistance of the upper part of the soil. 

9.3.2 Lateral Load Capacity of Pile 

In respect of the ultimate lateral resistance of a c'- φ' material, the method proposed for short rigid 

piles by Brinch Hansen (1911) can be referred (Figure 9.9). 

March 2009 9-31


Notes: 

1. The above passive pressure coefficients K qr and K cz are obtained based on the method 

proposed by Brinch Hansen (1961). Unit passive resistance per unit width, p z at depth z is: 

p z = σ v ’ K qz + c’K cz (9.17) 

where σ v ’ is the effective overburden pressure at depth z, c’ is the apparent cohesion of soil 

at depth z. 

9-32 March 2009


2. The point of rotation (Point X) is the point at which the sum of the moment (∑ M) of the 

passive pressure about the point of application of the horizontal load is zero. This point can 

be determined by a trial and adjustment process. 

∑ M= ∑ 

z=h L 

z=0 p z 

n e 1+zD- ∑ 

z=L L 

z=x p z 

n e 1+zD (9.18) 

3. The ultimate lateral resistance of a pile to the horizontal force H u can be obtained by taking 

moment about the point rotation, i.e. 

H u e 1 +x= ∑ 

z=h L 

z=0 p z 

n Dx-z- ∑ 

z=L 

p L 

z=x z n 

z-xD (9.19) 

4. An applied moment M can be replaced by a horizontal force H at a distance e 1 above the 

ground surface where M = H e 1 . 

5. When the head of a pile is fixed against rotation, the equivalent height, e o above the point of 

fixity of a force H acting on a pile with a free-head is given by e o = 0.5 (e 1 + z f ) is the depth 

from the ground surface to point of virtual fixity. ACI (1980) recommended that z f should be 

taken as 1.4R for stiff, overconsolidated clays and 1.8T for normally consolidatedclays, 

granular soils and silts, and peat. Pile stiffness factors, R and T, can ve determined based on 

Figure. 

Figure 9.8 Coefficients Kqz and Kcz at Depth z for Short Piles Subject to Lateral Load (Brinch Hansen, 

1911) 

Methods of calculating the ultimate lateral soil resistance for fixed-head and free-head piles in 

granular soils and clays are put forward by Broms (1914a & b). The theory is similar to that of Brinch 

Hansen except that some simplifications are made in respect of the distribution of ultimate soil 

resistance with depth. The design for short and long piles in granular soils are summarised in Figures 

9.10 and 9.11 respectively. Kulhawy & Chen (1992) compared the results of a number of field and 

laboratory tests on bored piles. They found that Brom’s method tended to underestimate the 

ultimate lateral load by about 15% to 20%. 

March 2009 9-33


Notes: 

1. For free-head short piles in granular soils 

2. 

H u = 0.5 DL3 K p s ' 

e 1 +L 

1+ sin ' 

Where K p = Rankine’s coefficient of passive pressure = 

1- sin ' 

D = width of the pile 

Ø’ = angle of shearing resistance of soil 

s = effective unit weight of soil 

3. For fixed-head short piles in granular soils 

4. H u = 1.5 DL 2 K P ρ C ’ 

The above equation is valid only when the maximum bending moment, M max 

develops at the pile head is less than the ultimate moment of resistance, M u , of the 

pile at this point. The bending moment is given by M max = DL 3 K P ρ C ’ 

5. P L is the concentrated horizontal force at pile tip due to passive soil resistance. 

Figure 9.9 Ultimate Lateral Resistance of Short Piles in Granular Soils (Broms, 1914a) 

9-34 March 2009


Notes: 

1. For free-head long piles in granular soils, M max = H(e 1 +0.67f*) 

where f* = 0.82 

H 

s 'DK p 

D = width of the pile in the direction of 

Ø’ = angle of shearing resistance 

s ’ = effective unit weight of soil 

= Rankine’s coefficient of passive 

K p 

2. For fixed-headed short piles in granular soils, the maximum bending moment 

occurs at the pile head and at the ultimate load. It is equal to the ultimate 

moment of resistance of pile shaft. 

M max = 0.5H (e 1 +0.67f*) 

For a pile of uniform cross-section, the ultimate value of lateral load H u is given 

by taking M max as the ultimate moment of resistance of the pile, M u . 

Figure 9.10 Ultimate Lateral Resistance of Long Piles in Granular Soils (Broms, 1914b) 

March 2009 9-35


Poulos (1985) has extended Broms' methods to consider the lateral load capacity of a pile in a twolayer 

soil. 

The design approaches presented above are simplified representations of the pile behaviour. 

Nevertheless, they form a useful framework for obtaining a rough estimate of the likely capacity, and 

experience suggests that they are generally adequate for routine design. 

In situations where the design is likely to be governed by lateral load behaviour, loading tests should 

be carried out to justify the design approach and verify the design parameters. The bending moment 

and shearing force in a pile subject to lateral loading may be assessed using the method by Matlock 

& Reese (1910) as given in Figures 9.12 and 9.13. The tabulated values of Matlock & Reese have 

been summarised by Elson (1984) for easy reference. This method models the pile as an elastic 

beam embedded in a homogeneous or non-homogeneous soil. 

In long, flexible piles, the structural capacity is likely to govern the ultimate capacity of a laterallyloaded 

pile. 

Relatively short less than critical length given in Figure 9.8 end-bearing piles, e.g. piles founded on 

rock, with toe being effectively fixed against both translation and rotation, can be modelled as 

cantilevers cast at the bottom, with the top either fixed or free, depending on restraints on pile head. 

Accordingly, the lateral stiffness of the overburden can thus be represented by springs with 

appropriate stiffness. 

9-36 March 2009


Deflection Coefficient, F s for Applied Moment M 

Deflection Coefficient, F s for Applied Lateral Load, H 

Moment Coefficient, F M for Applied Moment M 

Moment Coefficient, F M for Applied Lateral Load, H 

Shear Coefficient, F v for Applied Moment M 

Shear Coefficient, F v for Applied Lateral Load, H 

Notes: 

5 

1. T = E pI p 

where E p I p = bending stiffness of pile and n h = constant of 

n h 

horizontal subgrade reaction 

2. 

3. Obtain coefficients F δ ,F M and F v at appropriate depths desired and 

compute deflection, moment and shear respectively using the given 

formulae. 

Figure 9.11 Influence Coefficients for Piles with Applied Lateral Load and Moment (Flexible Cap or 

Hinged End Conditions) (Matlock & Reese, 1910) 

March 2009 9-37


Deflection Coefficient, F δ for Applied Lateral Load, H 

Moment Coefficient, F M for Applied Lateral Force, H 

Notes: 

5 

1. T = E pI p 

where E 

n p I p = bending stiffness of pile and n h = constant of horizontal 

h 

subgrade reaction 

2. Obtain coefficients F δ ,F M and F v at appropriate depths desired and compute 

deflection, moment and shear respectively using the given formulae. 

3. Maximum shear occurs at top of pile and is equal to the applied load H. 

Figure 9.12 Influence Coefficients for Piles with Applied Lateral Load (Fixed against Rotation at 

Ground Surface) (Matlock & Reese, 1910) 

9-38 March 2009


The minimum factors of safety recommended for design are summarised in Table 9.3. For vertical 

piles designed to resist lateral load, it is usually governed by the limiting lateral deflection 

requirements. 

For piles in sloping ground, the ultimate lateral resistance can be affected significantly if the piles are 

positioned within a distance of about five to seven pile diameters from the slope crest. Based on fullscale 

test results, Bhushan et al (1979) proposed that the lateral resistance for level ground be 

factored by 1/(1 + tan θ s ), where θ s is the slope angle. Alternatively, Siu (1992) proposed a 

simplifying method for determining the lateral resistance of a pile in sloping ground taking into 

account three-dimensional effects. 

9.3.3 Inclined Loads 

If a vertical pile is subjected to an inclined and eccentric load, the ultimate bearing capacity in the 

direction of the applied load is intermediate between that of a lateral load and a vertical load 

because the passive earth pressure is increased and the vertical bearing capacity is decreased by the 

inclination and eccentricity of the load. Based on model tests, Meyerhof (1981) suggested that the 

vertical component Q v , of the ultimate eccentric and inclined load can be expressed in terms of a 

reduction factor r f on the ultimate concentric vertical load Q o , as given in Figure 9.13. 

The lateral load capacity can be estimated following the methods given in Item 9.3.2 above. Piles, 

subjected to inclined loads, should also be checked against possible buckling, pile head deflection 

and induced bending moments. 

9.3.4 Raking Piles in Soil 

Raking piles provide a common method of resisting lateral loads. For the normal range of inclination 

of raking piles used in practice, the raking pile may be considered as an equivalent vertical pile 

subjected to inclined loading. 

Deformations and forces induced in a general pile group comprising vertical and raking piles under 

combined loading condition are not amenable to presentation in graphical or equation format. A 

detailed analysis will invariably require the use of a computer. 

Zhang et al (2002) conducted centrifuge tests to investigate the effect of vertical load on the lateral 

response of a pile group with raking piles. The results of the experiments indicated that there was a 

slight increase in the lateral resistance of the pile groups with the application of a vertical load. 

a) Methodologies for Analysis 

i) Stiffness method can be used to analyse pile groups comprising vertical piles and raking piles 

installed to any inclination. In this method, the piles and pile cap form a structural frame to carry 

axial, lateral and moment loading. The piles are assumed to be pin-jointed and deformed elastically. 

The load on each pile is determined based on the analysis of the structural frame. The lateral 

restraint of the soil is neglected and this model is not a good representation of the actual behaviour 

of the pile group. The design is inherently conservative and other forms of analyses are preferred for 

pile groups subjected to large lateral load and moment (Elson, 1984). 

ii) A more rational approach is to model the soil as an elastic continuum. A number of 

commercial computer programs have been written for general pile group analysis based on idealising 

the soil as a linear elastic material, e.g. PIGLET (Randolph, 1980), DEFPIG (Poulos, 1990a), PGROUP 

(Bannerjee & Driscoll, 1978). The first two programs are based on the interaction factor method 

March 2009 9-39


while the last one uses the boundary element method. A brief summary of the features of some of 

the computer programs developed for analysis of general pile groups can be found in Poulos (1989b) 

and the report by the Institution of Structural Engineers (ISE, 1989). Computer analyses based on 

the elastic continuum method generally allow more realistic boundary conditions, variation in pile 

stiffness and complex combined loading to be modelled. 

Comparisons between results of different computer programs for simple problems have been carried 

out, e.g. O'Neill & Ha (1982) and Poulos & Randolph (1983). The comparisons are generally 

favourable with discrepancies which are likely to be less than the margin of uncertainty associated 

with the input parameters. Comparisons of this kind lend confidence in the use of these programs for 

more complex problems. 

Pile group analysis programs can be useful to give an insight into the effects of interaction and to 

provide a sound basis for rational design decisions. In practice, however, the simplification of the 

elastic analyses, together with the assumptions made for the idealisation of the soil profile, soil 

properties and construction sequence could potentially lead to misleading results for a complex 

problem. Therefore, considerable care must be exercised in the interpretation of the results. 

The limitations of the computer programs must be understood and the idealisations and assumptions 

made in the analyses must be compatible with the problem being considered. It would be prudent to 

carry out parametric studies to investigate the sensitivity of the governing parameters for complex 

problems. 

b) Choice of Parameters 

One of the biggest problems faced by a designer is the choice of appropriate soil parameters for 

analysis. Given the differing assumptions and problem formulation between computer programs, 

somewhat different soil parameters may be required for different programs for a certain problem. 

The appropriate soil parameters should ideally be calibrated against a similar case history or derived 

from the back analysis of a site-specific instrumented pile test using the proposed computer program 

for a detailed analysis. 

9.3.5 Lateral Loading 


The response of piles to lateral loading is sensitive to soil properties near the ground surface. Due to 

the proneness to disturbance of these surface layers, reasonably conservative soil parameters should 

be adopted in the prediction of pile deflection. An approximate assessment of the effects of soil 

layering can be made by reference to the work by Davisson & Gill (1913) or Pise (1982). 

Poulos (1972) studied the behaviour of a laterally-loaded pile socketed in rock. He concluded that 

socketing of a pile has little influence on the horizontal deflection at working load unless the pile is 

sufficiently rigid, with a stiffness factor under lateral loading, K r , greater than 0.01, where 

K f = E pI p 

E s L 4 (9.20) 

and I p and L are the second moment of area and length of the pile respectively. 

The effect of sloping ground in front of a laterally-loaded pile was analysed by Poulos (1971) for 

clayey soils, and by Nakashima et al (1985) for granular soils. It was concluded that the effect on 

pile deformation will not be significant if the pile is beyond a distance of about five (5) to seven (7) 

pile diameters from the slope crest. 

9-40 March 2009


The load-deflection and load-rotation relationships for a laterally-loaded pile are generally highly 

non-linear. Three approaches have been proposed for predicting the behaviour of a single pile: 

(a) The equivalent cantilever method, 

(b) The subgrade reaction method, and 

(c) The elastic continuum method. 

Alternative methods include numerical methods such as the finite element and boundary element 

methods as discussed in the subsequent sections of this chapter. However, these are seldom 

justified for routine design problems. 

A useful summary of the methods of determining the horizontal soil stiffness is given by 

Jamiolkowski & Garassino (1977). 

It should be noted that the currently available analytical methods for assessing deformation of 

laterally-loaded piles do not consider the contribution of the side shear stiffness. Some allowance 

may be made for barrettes loaded in the direction of the long side of the section with the use of 

additional springs to model the shear stiffness and capacity in the subgrade reaction approach. 

Where the allowable deformation is relatively large, the effects of non-linear bending behaviour of 

the pile section due to progressive yielding and cracking, along with its effect on the deflection and 

bending moment profile should be considered (Kramer & Heavey, 1988). The possible non-linear 

structural behaviour of the section can be determined by measuring the response of an upstand 

above the ground surface in a lateral loading test. 

9.3.5.2 Equivalent Cantilever Method 

This method represents a gross simplification of the problem and should only be used as an 

approximate check on the other more rigorous methods unless the pile is subject to nominal lateral 

load. In this method, the pile is represented by an equivalent cantilever and the deflection is 

computed for either free-head or fixed-head conditions. Empirical expressions for the depths to the 

point of virtual fixity in different ground conditions are summarised by Tomlinson (1994). 

The principal shortcoming of this approach is that the relative pile-soil stiffness is not considered in a 

rational framework in determining the point of fixity. Also, the method is not suited for evaluating 

profiles of bending moments. 

9.3.5.3 Subgrade Reaction Method 

In this method, the soil is idealised as a series of discrete springs down the pile shaft. The continuum 

nature of the soil is not taken into account in this formulation. The characteristic of the soil spring is 

thus expressed as follows: 

p = k h δ h (9.21) 

P h = K h δ h (9.22) 

= k h D δ h (for constant K h ) 

= n h z δ h (for the case of K h varying linearly with depth) 

Where: 

p = soil pressure 

k h = coefficient of horizontal subgrade reaction 

δ h = lateral deflection 

March 2009 9-41


P h = soil reaction per unit length of pile 

K h = modulus horizontal subgrade reaction 

D = width or diameter of pile 

n h = constant of horizontal subgrade reaction, sometimes referred to as the constant of 

modulus variation in the literature 

z = depth below ground surface 

It should be noted that k h is not a fundamental soil parameter as it is influenced by the pile 

dimensions. In contrast, K h is more of a fundamental property and is related to the Young's modulus 

of the soil, and it is not a function of pile dimensions. Soil springs determined using subgrade 

reaction do not consider the interaction between adjoining springs. Calibration against field test data 

may be necessary in order to adjust the soil modulus to derive a better estimation (Poulos et al, 

2002). 

Traditionally, over-consolidated clay is assumed to have a constant K h with depth whereas normally 

consolidated clay and granular soil is assumed to have a K h increasing linearly with depth, starting 

from zero at ground surface. For a uniform pile with a given bending stiffness (E p I p ), there is a 

critical length (L c ) beyond which the pile behaves as if it were infinitely long and can be termed a 

'flexible' pile, under lateral load. 

The expressions for the critical lengths are thus given as follows: 

4 

L c = 4 E pI p 

K h 

(9.23) 

= 4 R for soils with a constant K h 

5 

L c = 4 E pI p 

n h 

= 4 T for soils with a K h increasing linearly with depth 

(9.24) 

The terms 'R' and 'T' are referred to as the characteristic lengths by Matlock & Reese (1910) for 

homogeneous soils and non-homogeneous soils, respectively. They derived generalised solutions for 

piles in granular soils and clayey soils. The solutions for granular soils as summarized in Figures 9.12 

and 9.13. 

A slightly different approach has been proposed by Broms (1914a & b) in which the pile response is 

related to the parameter L/R for clays, and to the parameter L/T for granular soils. The solutions 

provide the deflection and rotation at the head of rigid and flexible piles. 

In general, the subgrade reaction method can give satisfactory predictions of the deflection of a 

single pile provided that the subgrade reaction parameters are derived from established correlations 

or calibrated against similar case histories or loading test results. 

Typical ranges of values of n h , together with recommendations for design approach, are given in 

Table 9.5, previously. 

The parameter k h can be related to results of pressuremeter tests (CGS, 1992). The effects of pile 

width and shape on the deformation parameters are discussed by Siu (1992). 

The solutions by Matlock & Reese (1910) apply for idealised, single layer soil. The subgrade reaction 

method can be extended to include non-linear effects by defining the complete load transfer curves 

or 'p-y' curves. This formulation is more complex and a nonlinear analysis generally requires the use 

9-42 March 2009


of computer models similar to those described by Bowles (1992), which can be used to take into 

account variation of deformation characteristics with depth. In this approach, the pile is represented 

by a number of segments each supported by a spring, and the spring stiffness can be related to the 

deformation parameters by empirical correlations (e.g. SPT N values). Due allowance can and should 

be made for the strength of the upper, and often weaker, soils whose strength may be fully 

mobilised even at working load condition. 

Alternatively, the load-transfer curves can be determined based on instrumented pile loading tests, in 

which a series of 'p-y' curves are derived for various types of soils. Nip & Ng (2005) presented a 

simple method to back-analyse results of laterally loaded piles for deriving the 'p-y' curves for 

superficial deposits. Reese & Van Impe (2001) discussed factors that should be considered when 

formulating the 'p-y' curves. These include pile types and flexural stiffness, duration of loading, pile 

geometry and layout, effect of pile installation and ground conditions. 

Despite the complexities in developing the 'p-y' curves, the analytical method is simple once the nonlinear 

behaviours of the soils are modelled by the 'p-y' curves. This method is particularly suitable for 

layered soils. 

9.3.5.4 Elastic Continuum Method 

Solutions for deflection and rotation based on elastic continuum assumptions are summarised by 

Poulos & Davis (1980). Design charts are given for different slenderness ratios (L/D) and the 

dimensionless pile stiffness factors under lateral loading (K r ) for both friction and end-bearing piles. 

The concept of critical length is however not considered in this formulation as pointed out by Elson 

(1984). 

A comparison of these simplified elastic continuum solutions with those of the rigorous boundary 

element analyses have been carried out by Elson (1984). The comparison suggests that the solutions 

by Poulos & Davis (1980) generally give higher deflections and rotations at ground surface, 

particularly for piles in a soil with increasing stiffness with depth. 

The elastic analysis has been extended by Poulos & Davis (1980) to account for plastic yielding of 

soil near ground surface. In this approximate method, the limiting ultimate stress criteria as 

proposed by Broms (1915) have been adopted to determine factors for correction of the basic 

solution. 

An alternative approach is proposed by Randolph (1981b) who fitted empirical algebraic expressions 

to the results of finite element analyses for homogeneous and non-homogeneous linear elastic soils. 

In this formulation, the critical pile length, L c (beyond which the pile plays no part in the behaviour of 

the upper part) is defined as follows: 

2⁄ 

L c = 2r o E 7 

pe 

G c 

(9.25) 

Where: 

G * = G(1+0.75v s ) 

G c = mean value of G * over the critical length, L c , in a flexible pile 

G = shear modulus of soil 

r o = radius of an equivalent circular pile 

v s = Poisson’s ration of soil 

E p I p = bending stiffness of actual pile 

March 2009 9-43


E pe = equivalent Young’s modulus of the pile = 4E pI p 

r o 

4 

For a given problem, iterations will be necessary to evaluate the values of L c and G c . Expressions for 

deflection and rotation at ground level given by Randolph's elastic continuum formulation are 

summarised in Figure 9.14. 

Results of horizontal plate loading tests carried out from within a hand-dug caisson in completely 

weathered granite (Whiteside, 1981) indicate the following range of correlation: 

E h ' = 0.1 N to 1.9 N (MPa) (9.26) 

where E h ' is the drained horizontal Young's modulus of the soil. 

The modulus may be nearer the lower bound if disturbance due to pile excavation and stress relief is 

excessive. The reloading modulus was however found to be two to three times the above values. 

Plumbridge et al (2000b) carried out lateral loading tests on large-diameter bored piles and barrettes 

in fill and alluvial deposits. Testing arrangement on five sites included a 100 cycle bi-directional 

loading stage followed by a five-stage maintained lateral loading test. The cyclic loading indicated 

only a negligible degradation in pile-soil stiffness after the 100 cycle bi-direction loading. The 

deflection behaviour for piles in push or pull directions was generally similar. Based on the deflection 

profile of the single pile in maintained-load tests, the correlation between horizontal Young's 

modulus, E h ' and SPT N value was found to range between 3 N and 4 N (MPa). 

Lam et al (1991) reported results of horizontal Goodman Jack tests carried out from within a caisson 

in moderately to slightly (Grade III / II) weathered granite. The interpreted rock mass modulus was 

in the range of 3.1 to 8.2 GPa. 

In the absence of site-specific field data, the above range of values may be used in preliminary 

design of piles subject to lateral loads. 

Free-head Piles 

δ h = E p/G c 1 

7 

0.27H 

+ 0.3M 

ρ c 'G c 0.5L c 0.5L c 2 

Ө = E p/G c 1 

7 

0.3H 

+ 0.8 'M 

 

ρ c 'G c . 0.5L c 3 

The maximum moment for a pile under a lateral load H occurs 

at depth between 0.25L c (for homogenous soil) and 0.33L c 

(for soil with stiffness proportional to depth). The value of the 

maximum bending moment M max may be approximated using 

the following expression: 

M max = 0.1 H L c 

ρ c ' 

Figure 9.13 Analysis of Behaviour of a Laterally Loaded Pile Using the Elastic Continuum Method 

(Randolph, 1981a) 

9-44 March 2009


In this case, the pile rotation at ground surface, Ө, equals zero and the fixing moment, M f , and 

lateral deflection, δ h , are given by the following expression: 

M f = - 0.375H(o.5L c) 

ρ c ' 

(9.27) 

δ h = (E p/G c ) 1 

7 

0.27- 0.11 

 

H 

(9.28) 

ρ c 'G c ρ c ' 

0.5L c 

Where: 

δ h 

Ө 

G c 

L c 

E po 

= lateral pile deflection at ground surface 

= pile rotation at ground surface 

= characteristic shear modulus, i.e. average value of G* over the critical length L c of 

the pile 

 

E 

= critical pile length for lateral loading = 2r o 

G 

= equivalent Young’s modulus of pile = 4E pI p 

r o 

4 

c ’ = degree of homogeneity over critical length, L c = G* 0.25Lc 

G c 

G* = G(1+0.75v s ) 

G* 0.25Lc = value of G* at depth of 0.25L c 

v c = Poisson’s ratio of soil 

G = shear modulus of soil 

H = horizontal load 

M = bending moment 

E p I p = bending stiffness of pile 

= pile radius 

r o 

The lateral deflection of a fixed-head pile is approximately half that of a corresponding free-head 

pile. 

9.4 PILE GROUP 

9.4.1 General 

Piles installed in a group to form a foundation will, when loaded, give rise to interaction between 

individual piles as well as between the structure and the piles. The pile- soil-pile interaction arises 

as a result of overlapping of stress (or strain) fields and could affect both the capacity and the 

settlement of the piles. The piled foundation as a whole also interacts with the structure by virtue 

of the difference in stiffness. This foundation-structure interaction affects the distribution of loads 

in the piles, together with forces and movements experienced by the structure. 

The analysis of the behaviour of a pile group is a complex soil-structure interaction problem. 

The behaviour of a pile group foundation will be influenced by, inter alia: 

(a) Method of pile installation, e.g. replacement or displacement piles, 

(b) Dominant mode of load transfer, i.e. shaft resistance or end- bearing, 

(c) Nature of founding materials, 

(d) Three-dimensional geometry of the pile group configuration, 

March 2009 9-45


(e) Presence or otherwise of a ground-bearing cap, and 

(f) Relative stiffness of the structure, the piles and the ground. 

Traditionally, the assessment of group effects is based on some 'rules-of-thumb' or semiempirical 

rules derived from field observations. Recent advances in analytical studies have 

enabled more rational design principles to be developed. With improved computing capabilities, 

general pile groups with a combination of vertical and raking piles subjected to complex loading 

can be analysed in a fairly rigorous manner and parametric studies can be carried out relatively 

efficiently and economically. 

9.4.2 Minimum Spacing of Piles 

The minimum spacing between piles in a group should be chosen in relation to the method of 

pile construction and the mode of load transfer. It is recommended that the following 

guidelines on minimum pile spacing may be adopted for routine design: 

(a) For bored piles which derive their capacities mainly from shaft resistance and for all types 

of driven piles, minimum centre-to-centre spacing should be greater than the perimeter of the pile 

(which should be taken as that of the larger pile where piles of different sizes are used); this 

spacing should not be less than 1 m as stipulated in the Code of Practice for Foundations (BD, 

2004a). 

(b) For bored piles which derive their capacities mainly from end-bearing, minimum clear spacing 

between the surfaces of adjacent piles should be based on practical considerations of positional and 

verticality tolerances of piles. It is prudent to provide a nominal minimum clear spacing of about 

0.5 m between shaft surfaces or edge of bell-outs. For mini-piles socketed into rock, the minimum 

spacing should be taken as the greater of 0.75 m or twice the pile diameter (BD, 2004a). 

The recommended tolerances of installed piles are shown in Table 9.6 (HKG, 1992). Closer 

spacing than that given above may be adopted only when it has been justified by detailed 

analyses of the effect on the settlement and bearing capacity of the pile group. Particular 

note should be taken of adjacent piles founded at different levels, in which case the effects of the 

load transfer and soil deformations arising from the piles at a higher level on those at a lower 

level need to be examined. The designer should also specify a pile installation sequence within a 

group that will assure maximum spacing between shafts being installed and those recently 

concreted. 

Table 9.6 Tolerance of Installed Piles 

Description 

Tolerance 

Land Piles Marine Piles 

Deviation from specified position in plan, 

measured at cut-off level 

75 mm 150 mm 

Deviation from vertical 1 in 75 1 in 25 

Deviation of raking piles from specified batter 

Deviation from specified cut-off level 

1 in 25 

25 mm 

The diameter of cast in-place piles shall be at least 97% of the specified diameter 

9.4.3 Ultimate Capacity of Pile Groups 

Traditionally, the ultimate load capacity of a pile group is related to the sum of ultimate capacity of 

individual piles through a group efficiency (or reduction) factor, η, defined as follows: 

9-46 March 2009


 

ultimate load capacity of a pile group 

sum of ultimate load capacities of individual piles in the group 

(9.29) 

A number of empirical formulae have been proposed, generally relating the group efficiency 

factor to the number and spacing of piles. However, most of these formulae give no more than 

arbitrary factors in an attempt to limit the potential pile group settlement. A comparison of a 

range of formulae made by Chellis (1911) shows a considerable variation in the values of η for a 

given pile group configuration. 

There is a lack of sound theoretical basic on the rationale and field data in support of the 

proposed empirical formulae (Fleming & Thorburn 1983). The use of these formulae to calculate 

group efficiency factors is therefore not recommended 

A more rational approach in assessing pile group capacities is to consider the capacity of both the 

individual piles (with allowance for pile-soil-pile interaction effects) and the capacity of the 

group as a block or a row and determine which failure mode is more critical. There must be an 

adequate factor of safety against the most critical mode of failure. 

The degree of pile-soil-pile interaction, which affects pile group capacities, is influenced by the 

method of pile installation, mechanism of load transfer and nature of the founding materials. The 

group efficiency factor may be assessed on the basis of observations made in instrumented model 

and field tests as described below. Generally, group interaction does not need to be considered 

where the spacing is in excess of about eight pile diameters (CGS, 1992). 

March 2009 9-47


REFERENCES 

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Report ACI 5438-74. American Concrete Institute. 

[2] ASTM (1995b). Standard Test Method for High-Strain Dynamic Testing Of Piles, D 4945-89. 

1995 Annual Book of ASTM Standards, vol. 04.09, American Society for Testing and Materials, New 

York, pp 10-11. 

[3] Bannerjee, P.K. & Driscoll, R.M.C. (1978). Program For The Analysis Of Pile Groups Of Any 

Geometry Subjected To Horizontal And Vertical Loads And Moments, PGROUP. HECB/B/7, 

Department of Transport, HECB, London, 188 p. 

[4] Bhushan, K., Haley, S.C. & Fong, P.T. “Lateral Load Tests on Drilled Piers in Stiff Clays.” 

Journal of the Geotechnical Engineering Division, American Society of Civil Engineers, vol. 105, pp 

919-985, 1979. 

[5] Bjerrum, L. & Eggestad, A. “Interpretation of Loading Test on Sand.” Proceedings of 

European Conference in Soil Mechanics, Wiesbaden, 1, pp 199-203, 1913. 


New York, 1992, 1004 p. 


New York, 1992, 1004 p. 

[8] Brinch Hansen, J. “The ultimate resistance of rigid piles against transversal forces. Danish 

Geotechnical Institute Bulletin, no. 12, pp 5-9. 1961 

[9] Broms, B.B. “The lateral resistance of piles in cohesive soils.” Journal of the Soil Mechanics 

and Foundations Division, American Society of Civil Engineers, vol. 90, no. SM2, pp 27-13, 1914a. 

[10] Broms, B.B. “The lateral resistance of piles in cohesionless soils.” Journal of the Soil 

Mechanics and Foundations Division, American Society of Civil Engineers, vol. 90, no. SM3, pp 123- 

151, 1914b. 

[11] Broms, B.B. “Design of laterally loaded piles.” Journal of the Soil Mechanics and Foundations 

Division, American Society of Civil Engineers, vol. 91, no. SM3, pp 79-99, 1915. 

[12] BSI. Eurocode 7: Geotechnical Design – Part 3: Design Assisted by Field Testing (DD ENV 

1997-3:2000). British Standards Institution, London, 2000b, 141 p. 



[14] Buisman, A.S.K. “Results of long duration settlement tests.” Proceedings of the First 


vol. 1, pp 103-101, 1931. 

[15] Burland, J.B. & Burbidge, M.C. “Settlement of foundations on sand and gravel.” Proceedings 

of Institution of Civil Engineers, Part 1, vol. 78, pp 1325-1381, 1985. 

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[16] GEO, Guide to Retaining Wall Design (Geoguide 1). (Second edition). Geotechnical 

Engineering Office, Hong Kong, 1993, 217 p. 

[17] CGS. Canadian Foundation Engineering Manual. (Third edition). Canadian Geotechnical 


[18] Chan, H.F.C. & Weeks, R.C. “Electrolevels or servo-accelerometers?’ Proceedings of the 

Fifteen Annual Seminar, Geotechnical Division, Hong Kong Institution of Engineers, pp 97-105, 1995. 

[19] Davisson, M.T. & Gill, H.L. ”Laterally loaded piles in a layered soil system.” Journal of the Soil 

Mechanics and Foundations Division, American Society of Civil Engineers, vol. 89, no. SM3, pp 13-94, 

1913. 

[20] Duncan, J. M., Buchignani, A. L., and DWet, M., An Engineering Manual for Slope Stability 

Studies, Department of Civil Engineering, Geotechnical Engineering, Virginia Polytechnic Institute 

and State University, Blacksburg, VA, 1987. 



[22] Elson, W.K. (1984). Design of Laterally-loaded Piles (CIRIA Report No. 103). Construction 

Industry Research & Information Association, London, 81 p. 

[23] EM 1110-2-1902. “Engineering and Design of Slope Stability,” U.S. Army Corp of Engineer, 


[24] EM 1110-2-1913. “Design and Construction of Levees,” U.S. Army Corp of Engineer, 


[25] Fraser, R.A. & Ng, H.Y. (1990). Pile failure. Proceedings of the Ninth Annual Seminar on 

Failures in Geotechnical Engineering, Geotechnical Division, Hong Kong Institution of Engineers, 

Hong Kong, pp 75-94 


Press, 374 p. 

[27] GCO (1984).” Geotechnical Manual for Slope”. (Second Edition). Geotechnical Control Office, 

Hong Kong 

[28] GCO (1990) “Review of Design Method for Excavation”. Geotechnical Control Office, Hong 

Kong 





[31] Jamiolkowski, M. & Garassino, A. (1977). Soil modulus for laterally loaded piles. Proceedings 

of the Specialty Session on the Effect of Horizontal Loads on Piles due to Surcharge or Seismic 

Effects, Ninth International Conference on Soil Mechanics and Foundation Engineering, Tokyo, pp 43- 

58. 

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[32] Kramer, S.L. & Heavey, E.J. (1988). Lateral load analysis of non-linear piles. Journal of 

Geotechnical Engineering, American Society of Civil Engineers, vol. 114, pp 1045-1049. 

[33] Kulhawy, F.H. & Chen, Y.J. (1992). A thirty-year perspective of Broms' lateral loading 

models, as applied to drilled shaft. Proceedings of the Bengt B. Broms Symposium on Geotechnical 

Engineering, Singapore, pp 225-240. 

[34] Lam, T.S.K., Tse, S.H., Cheung, C.K. & Lo, A.K.Y. (1994). Performance of two steel Hpiles 

founded in weathered meta-siltstone. Proceedings of the Fifth International Conference on Piling and 

Deep Foundations, Brugge, pp 5.1.1-5.1.10. 

[35] Lam, T.S.K., Yau, J.H.W. & Premchitt, J. (1991). Side resistance of a rock-socketed caisson. 

Hong Kong Engineer, vol. 19, no. 2, pp 17-28. 

[36] Matlock, H. & Reese, L.C. (1910). Generalised solutions for laterally-loaded piles. Journal of 

the Soil Mechanics and Foundations Division, American Society of Civil Engineers, vol. 81, no. SM3, 

pp 13-91. 



[38] Meyerhof, G.G. (1981). Theory and practice of pile foundations. Proceedings of the 

International Conference on Deep Foundations, Beijing, vol. 2, pp 1.77-1.81. 

[39] Nakashima, E., Tabara, K. & Maeda, Y.C. (1985). Theory and design of foundations on 

slopes. Proceedings of Japan Society of Civil Engineers, no. 355, pp 41-52. (In Japanese). 

[40] Ng, H.Y.F. (1989). Study of the Skin Friction of a Large Displacement Pile. M.Sc. Dissertation, 

University of Hong Kong, 200 p. (Unpublished). 

[41] Nip, D.C.N. & Ng, C.W.W (2005). Back-analysis of laterally loaded piles. Proceedings of the 

Institution of Civil Engineers, Geotechnical Engineering, vol. 158, pp 13 - 73. 

[42] O'Neill, M.W. & Ha, H.B. (1982). Comparative modelling of vertical pile groups. Proceedings 

of the Second International Conference on Numerical Methods in Offshore Piling, Austin, pp 399-418. 

[43] O'Neill, M.W. & Reese, L.C. (1999). Drilled Shaft : Construction Procedures and Design 

Methods. Federal Highway Administration, United States, 790 p. 




[45] Pise, P.J. (1982). Laterally loaded piles in a two-layer soil system. Journal of Geotechnical 

Engineering, American Society of Civil Engineers, vol. 108, pp 1177-1181. 

[46] Plumbridge, G.D., Sze, J.W.C. & Tham, T.T.F. (2000b). Full-scale lateral load tests on bored 

piles and a barrette. Proceedings of the Nineteenth Annual Seminar, Geotechnical Division, Hong 

Kong Institution of Engineers, pp 211-220. 

[47] Poulos, H.G. & Davis, E.H. (1974). Elastic Solutions for Soil and Rock Mechanics. John Wiley 

& Sons, New York, 411 p. 

9-50 March 2009


[48] Poulos, H.G. & Davis, E.H. (1980). Pile Foundation Analysis and Design. John Wiley & Sons, 

New York, 397 p. 

[49] Poulos, H.G. & Randolph, M.F. (1983). Pile group analysis: a study of two methods. Journal 

of Geotechnical Engineering, American Society of Civil Engineers, vol. 109, pp 355-372. 

[50] Poulos, H.G. (1972). Behaviour of laterally loaded piles: III - socketed piles. Journal of the 

Soil Mechanics and Foundations Division, American Society of Civil Engineers, vol. 98, pp 341-311. 

[51] Poulos, H.G. (1971). Behaviour of laterally loaded piles near a cut slope. Australian 

Geomechanics Journal, vol. G1, no. 1, pp 1-12. 

[52] Poulos, H.G. (1985). Ultimate lateral pile capacity in a two-layer soil. Geotechnical 

Engineering, vol. 11, no. 1, pp 25-37. 

[53] Poulos, H.G. (1989b). Pile behaviour - theory and application. Géotechnique, vol. 39, pp 315- 

415. 

[54] Poulos, H.G. (1990a). DEFPIG Users' Manual. Centre for Geotechnical Research, University of 

Sydney, 55 p. 

[55] Poulos, H.G. (2000). Foundation Settlement Analysis – Practice versus Research. The Eighth 

Spencer J Buchanan Lecture, Texas, 34 p. 






pp 2.13-2.72. 

[58] Randolph, M.F. (1980). PIGLET: A Computer Program for the Analysis and Design of Pile 

Groups under General Loading Conditions (Cambridge University Engineering Department Research 

Report, Soils TR 91). 33 p. 

[59] Randolph, M.F. (1981b). The response of flexible piles to lateral loading. Géotechnique, vol. 

31, pp 247-259. 

[60] Reese, L.C. & Van Impe, W.F. (2001). Single Piles and Pile Group under Lateral Loading. 

Rotterdam, Balkema, 413 p. 



[62] Siu, K.L. (1992). Review of design approaches for laterally-loaded caissons for building 

structures on soil slopes. Proceedings of the Twelfth Annual Seminar, Geotechnical Division, Hong 

Kong Institution of Engineers, Hong Kong, pp 17-89. 

[63] Smith, E.A.L. (1912). Pile-driving analysis by the wave equation. Transactions of the 

American Society of Civil Engineers, vol. 127, pp 1145-1193. 

March 2009 9-51





297-321. 

[66] Tomlinson, M.J. (1994). Pile Design and Construction Practice. (Fourth edition). Spon, 411 p. 



[68] Weltman, A.J. (1980b). Pile Load Testing Procedures (CIRIA Report No. PG7). Construction 

Industry Research & Information Association, London, 53 p. 

[69] Whiteside, P.G. (1981). Horizontal plate loading tests in completely decomposed granite. 

Hong Kong Engineer, vol. 14, no. 10, pp 7-14. 

[70] Yoshida, I. & Yoshinaka, R. (1972). A method to estimate soil modulus of horizontal 

subgrade reaction for a pile. Soils and Foundations, vol. 12(3), pp 1-11. 

[71] Zhang, L.M., McVay, M.C., Han, S.J., Lai, P.W. & Gardner, R. (2002). Effect of dead loads on 

the lateral response of battered pile groups. Canadian Geotechnical Journal, vol. 39, pp 511-575. 

9-52 March 2009

CHAPTER 10 SEEPAGE

Chapter 10 SEEPAGE 



List of Tables ........................................................................................................................10-ii 

List of Figures .......................................................................................................................10-ii 

10.1 SEEPAGE .................................................................................................................. 10-1 

10.2 LANE’S WEIGHTED CREEP THEORY ............................................................................. 10-1 

10.3 FLOWNETS ............................................................................................................... 10-3 

10.4 CONTROL OF SEEPAGE .............................................................................................. 10-5 

10.5 PROTECTIVE FILTER REQUIREMENTS ......................................................................... 10-5 

REFERENCES ....................................................................................................................... 10-7 

March 2009 10-i




10.1 Lane’s Weighted-Creep Ratios 10-1 

10.2 Gradation Requirements For Filter Materials (after USBR, 1974) 10-6 


‘ 


10.1 Example of Application of Lane’s Weighted Creep Theory on a Dam on Pervious 

Foundation 10-2 

10.2 Flownet Illustrating Some Definitions 10-3 

10.3 Example Calculation - Flownet 10-4 

10-ii March 2009


10 SEEPAGE 

10.1 SEEPAGE 

When water flows through a porous medium such as soil, energy or head is lost through friction 

similar to what happens in flow through pipes and open channels. For example, energy or head 

losses occur when water seeps through an earth dam or under a sheet pile cofferdam (Figure 10.1 

(a) and (b)). The flow through the soils also exert seepage forces on the individual soil grains, 

which affect the intergranular or effective stresses in the soil masses. Seepage can create problems 

especially in water control structures such as excessive seepage losses, uplift pressures and 

potential detrimental piping and erosion. 

This section discusses two of the many methods available which are simple and easy to use. They 

are Lane’s weighted creep theory and flownets. Flownets, if properly constructed are more 

accurate than the former and result in more realistic determinations of seepage pressure and piping 

potential. 

10.2 LANE’S WEIGHTED CREEP THEORY 

Lane’s theory may be used for designing low concrete hydraulic structures on pervious foundations. 

The concept is based on the following principles:- 

a) The weighted-creep distance of a cross section of a hydraulic structure is the sum of the 

vertical creep distances (steeper than 45°) plus one-third of the horizontal creep distances 

(less than 45°). 

b) The weighted-creep head ratio is the weighted-creep distance divided by the effective head. 

c) Reverse filter drains, weep holes, and pipe drains are aids to security from underseepage, 

and recommended safe weighted-creep head ratios may be reduced as much as 10 percent 

if they are used. 

d) Care must be exercised to ensure that cutoffs are properly tied in at the ends so that the 

water will not outflank them. 

e) The upward pressure to be used in the design may be estimated by assuming that the drop 

in pressure from headwater to tailwater along the contact line of the hydraulic struicture and 

the foundation is proportional to the weighted-creep distance. 

The Lane’s weighted-creep ratios are as shown in Table 10.1. 

Table 10.1 Lane’s Weighted-Creep Ratios 

Materials 

Ratio 

Very fiine sand or silt 8.5 

Fine sand 7.0 

Medium sand 6.0 

Coarse sand 5.0 

Fine gravel 4.0 

Medium gravel 3.5 

Coarse gravel including cobbles 3.0 

Boulders with some gravels and conbbles 2.5 

Soft clay 3.0 

Medium clay 2.0 

Hard clay 1.8 

Very hard clay or hardpan 1.6 

March 2009 10-1


Figure 10.1 is an example of the application of Lane’s Weighted Creep Theoy for the design of a 

concrete dam or spillway. This example determines the magnitude of uplift pressures at various 

points under the structure and any potential piping problem for the headwater and tailwater 

conditions shown. 

Normal water surface (headwater) 

Upstream Apron 

4.5 m 

7.5 m 

Point A 

Downstream Apron 

1.0 m 1.0 m 

Point B 

10 m 10 m 

10 m 

Tailwater surface 

1.5 m 

Figure 10.1 Example of Application of Lane’s Weighted Creep Theory on a Dam on Pervious 

Foundation 

Weighted length of path = 4.5 + 4.5 + (4 x 1) + 1/3 (10 + 10 + 10) = 23 m. 

Head on structure = Headwater – tailwater = 7.5 – 1.5 = 6 m 

Weighted – creep ratio = 23 = 3.83 

6 

According to Lane’s recommended ratios, this dam would be safe from piping on clay or on medium 

and coarse gravel, but not on silt, sand, or fine gravel. With properly placed drains and filters, the 

structure would probably be considered safe on a fine gravel foundation as discussed in Item 10.2 

principle (c). 

Uplift, point A = ( 7.5 – 1.5 ) - (4.5 + 4.5 + 10/3) x 6 

23 

+ 1.5 (depth of tailwater above foundation level) 

= 6 – 4.61 + 1.5 

= 4.28 m 

Uplift, point B = (7.5 – 1.5) - (4.5 + 4.5 + 10/3 + 1 + 1 + 10/3) x 6 + 1.5 

23 

= 6 – 4.61 + 1.5 

= 2.9 m 

Total Uplift = 

(4.28 + 2.9) x 9.81 x 10 

2 

= 352.2 N per m of crest length of dam. 

The weighted-creep head ratio can be increased by increasing the depth of the upstream cutoff or 

by increasing the apron length. Either of these alternative would also decrease the uplift under the 

structure. 

10-2 March 2009

Chapter 

10 SEEPAGE 

10.3 

FLOWNETS 

The flow 

of water through a soil can be represented graphically by means of a flownet, whichh 

consists of flow lines and equipotential lines. 

Flow Lines 

The paths that the water follows in 

the course of seepage are known as flow lines. 

Equipotential Lines 

As the water moves along the flow line, it experiences a continuous loss of head. If the head 

causing flow at points along a flow 

line can be 

obtained, then by joining up points of equal head 

potential, a second set of lines known as equipotential lines are obtained. Hence, along an 

equalpotential line, the energy available to cause flow is the same; conversly, the energy loss by 

the water getting to that line is the 

same all along that line. 

If from the infinite number of flow 

lines possible we choose 

only a few in such a manner that the 

same fraction ∆q of the total seepage is passing between any pair of neighbouring flow lines, and 

similarly, 

if we choose 

from the infinite number of possible equipotential lines only a few in such a 

manner that the drop in head ∆h between any 

pair of neighbouring equipotential lines is equal to a 

constant fraction of the total loss in head h, then the resulting flow net possesses the 

property that 

the ratio 

of the sides of each rectangle, bordered by two flow and two equipotential lines, is 

constant. If all sidess of one such rectangle are equal, then the entiree flow net must consist of 

squares. If one succeeds in plotting two sets of curves so that they 

intersect at 

right angles, 

forming squares and fulfilling boundary conditions, then one 

has solved graphically the problems of 

seepage. 

Figure 10.2 Flownet Illustrating Some Definitions 

From the 

flownet, the 

designer may gather:- 

a) Uplift forces 

b) Exit hydraulic gradients (which is a measure of piping potential) and 

c) Quantity of seepage 

The following gives an example of a seepagee problem solved by means of flownet 

for the case 

where the permeability of the soil is isotropic i.e. horizontal permeability equals 

the vertical 

permeability. The factor of safety required against piping is normally greater than 3.0 

March 2009 

10-3

Chapter 

10 SEEPAGE 

Figure 

10.3 Example Calculation 

- Flownet 

No flow channels, N f = 4 

No pressure drops, N d = 10 

(note feet of head acting at each equipotential). 

∆h = = 

= 

Seepage 

q = kH = 

= (0.00305 m/min) (6 m) (4/10) (1 m wide) 

= 7.32 x 10 

-3 m/min per meter width. 

Uplift Force on Base 

L = 14.5m 

p A = (1.5 m + 2.1 m) 

p B = (1.5 m + 0.6 m) 

Uplift = (L) = 

= 405 kN per meter of width 

Escape Gradient at Downstream Tip 

∆h between last two equipotential lines = 0.6 m 

L = 1.83 m 

I = - 2 = = 0.33 

Critical exit gradient i crit = 1 

p A 

p B 

10-4 

March 2009


Factor of safety against piping = i crit 

i 

= 3 

The above example assumed the permeability of the soil to be isotropic. Generally, the horizontal 

(K h ) and vertical coefficients of permeability (K v ) of a soil differ, usually the former is greater than 

the latter. In such instances, the method of drawing the flownet need to be modified. Use of a 

transformed section is an easily applied method which accounts for the different rates of 

permeability. 

Vertical dimensions are selected in accord with the scale desired for the drawing. Horizontal 

dimensions, however, are modified by multiplying all horizontal lengths by the factor √(k v /k h ). The 

conventional flownet is then drawn on the transformed section. For flow through the anisotropic 

soil, the seepage, q is 

q=H w 

H w 

N f 

N d 

N f 

N d 

K v K h (10.1) 

= head difference 

= number of flow channels 

= number of pressure drops 

In addition to the flow net and weighted-creep methods of estimating the distribution of uplift 

pressure are Khosla’s method of independent variables and Rao’s relaxation method which can be 

used for making computations of uplift at critical points along the base of the structure. Because 

these theories are highly mathematical they are not discussed in this text. 

10.4 CONTROL OF SEEPAGE 

Piping can occur any place in the system, but usaully it occurs where the flow is concentrated e.g. 

at the downstream toe of the dam or at any place where seepage water exits. Once seepage forces 

are large enough to move particles, piping and erosion can start, and usually continues until either 

all the soils in the vicinity are carried away or the structure collapses. Cohesionless soils, especially 

silty soils, are highly susceptible to piping 

Uplift and seepage problems may be alleviated or controlled by several methods. Among which are: 

a) Construction of cut-off wall or trench to completely block the seeping water 

b) Installation of an impervious blanket e.g an apron to lengthen the drainage path so that 

more of the head is lost and thus the hydraulic gradient in the critical region is reduced. 

c) Installation of relief wells and other kinds of drains can be used to relief high uplift 

pressures at the base of hydraulic structures 

d) Installation of protective filter, which consists of one or more layers of free-draining 

granular materials placed in less pervious foundation or base materials to prevent the 

movement of soil particles that are susceptible to piping while at the same allowing the 

seepage water ro escape with relatively little head loss. The requirements for a protective 

filter are discussed in Item 10.5 below 

10.5 PROTECTIVE FILTER REQUIREMENTS 

In generaal, the four basic requirements of the protective filter layer for controlling the seepage 

problems such as piping and uplift pressures are as follows: 

March 2009 10-5


a) The filter material should be more pervious than the base material in order that no 

hydraulic pressure will build up to disrupt the filter and adjacent structures 

b) The voids of the inplace filter material must be small enough to prevent base material 

particles from penetrating the filter and causing clogging and failure of the protective filter 

system. 

c) The layer of the protective filter must be sufficiently thick to provide a good distribution of 

all particles sizes throughout the filter 

d) Filter material particles must be prevented from movement into the drainage pipes by 

providing sufficientlyy small slot openings or perforations or additional coarser filter zones if 

necessary. This requirement could also be fulfilled by using some of the non-woven and 

woven fabric materials developed recently. 

The gradation requirements for protective filters are given in Table 10.2. The first ratio, R 15 , 

ensures that the small particles of the material to be protected are prevented from passing through 

the pores of the filters; the second ratio, R 50 , ensures that the seepage forces witin the filter are 

reasonably small. If the criteria in this table cannot be met by one layer of filter material, then a 

zoned or multilayered filter can be designed and specified. 

Table 10.2 Gradation Requirements For Filter Materials (after USBR, 1974) 

Filter Materials Characteristics R 15 R 50 

Uniform grain size filters, C u = 3 to 4 - 5 to 10 

Graded filters, subrounded particles 12 to 40 12 to 58 

Graded filters, angular particles 6 to 18 9 to 30 

R15 = D15 of filter material 

D15 of material to be protected 

R50 = 

D50 of filter material 

D50 of material to be protected 

Notes: 

Maximum size of the filter material should be less than 76 mm. Use the 

minus No. 4 fraction of the base material for setting filter limits when 

the gravel content (plus No. 4) is more than 10%, and the fines (minus 

No. 200) are more than 10%. Filters must not have more than 5% 

minus No. 200 particles to prevent excessive movement of fines in the 

filter and into drainage pipes. The grain size distribution curves of the 

filter and the base material should approximately parallel in the range of 

finer sizes. 

10-6 March 2009


REFERENCES 


New York, 1992, 1004 p. 











[8] GCO (1984). Geotechnical Manual for Slope. (Second Edition). Geotechnical Control Office, 

Hong Kong 


Kong 



[11] Harry R.Cedergreen, Seepage, Drainage and Flownet, John Wiley nd Sons. 

[12] Heerten G., Dimensioning the filtration properties of geotextiles considering long term 

conditions, Proceedings 2nd. International Conference on Geotextiles, Las Vegas, Vol.1, pp. 115 - 

120. 


New Jersey 


[15] Lane, E.W., Security from Underseepage, Tran. ASCE, Vol. 100, 1935 p.1235. 

[16] Lawson C.R., Geotextiles, Unpublished. 

[17] Lawson C.R., Filter Criteria for Geotextiles Relevance and Use" Journal of Geotechnical 

Engineering Division ASCE. Vol. lO8, GT10, 1982. 



1974. 

March 2009 10-7






[23] United Bureau States Department of the Interior, Design of Small Dams Bureau of 


10-8 March 2009































March 2009 

i



Amend 

No 

Page 

No 

Date of 

Amendment 

Amend 

No 

Page 

No 

Date of 

Admendment 

ii March 2009






Chapter 1 

Chapter 2 

Chapter 3 

Chapter 4 

Chapter 5 

PLANNING AND SCOPE 

SAMPLING AND SAMPLING DISTURBANCE 

IN SITU GEOTECHNICAL TESTING 

LAB TESTING FOR SOILS 

INTERPRETATION OF SOIL PROPERTIES 

March 2009 

iii



iv March 2009

PART 2: SOIL INVESTIGATION

CHAPTER 1 PLANNING AND SCOPE

Chapter 1 PLANNING AND SCOPE 


Table of Contents ................................................................................................................... 1-i 

List of Table ........................................................................................................................... 1-ii 


1.1 INTRODUCTION .......................................................................................................... 1-1 

1.2 GENERAL .................................................................................................................... 1-1 

1.3 OBJECTIVES ................................................................................................................ 1-1 

1.4 PHASES OF INVESTIGATIONS ...................................................................................... 1-2 

1.5 APPROACHES TO SITE INVESTIGATIONS ...................................................................... 1-3 

1.5.1 Approach 1: Reconnaissance – Site Visit ......................................................... 1-3 

1.5.2 Approach 2: Desk-Study and Geotechnical Advice ............................................ 1-3 

1.5.3 Approach 3: Ground Investigation .................................................................. 1-4 

1.6 EXPLORATION AND SAMPLING ..................................................................................... 1-5 

1.6.1 Spacing of Pits and Borings ............................................................................ 1-6 

1.6.2 Depths of Borings ......................................................................................... 1-9 

1.6.3 Sampling, Laboratory Testing and In situ Testing Requirements ....................... 1-12 

1.7 METHODS OF SITE INVESTIGATION – DRILLING AND SAMPLING .................................. 1-17 

1.7.1 Subsurface Exploration ................................................................................. 1-17 

1.7.2 Boring ......................................................................................................... 1-18 

1.7.2.1 Light Percussion Drilling ............................................................ 1-18 

1.7.2.2 Augering.................................................................................. 1-19 

1.7.2.3 Wash Boring ............................................................................ 1-20 

1.7.3 Drilling ........................................................................................................ 1-21 

1.7.3.1 Open-Holing ............................................................................ 1-21 

1.7.3.2 Coring ..................................................................................... 1-21 

1.7.4 Exploration Pit Excavation ............................................................................. 1-24 

1.7.5 Probing ....................................................................................................... 1-24 

1.7.5.1 MacKintosh Probe ..................................................................... 1-24 

1.7.6 Examination In-Situ ...................................................................................... 1-25 

1.7.6.1 Trial Pit ................................................................................... 1-25 

REFERENCES ....................................................................................................................... 1-27 

March 2009 1-i


List of Table 


1.1 Planning a Ground Investigation 1-6 

1.2 Recommended Number and Depth of Borings 1-7 

1.3 Relative Merits of In Situ and Laboratory Testing 1-14 

1.4 Common Uses of In Situ and Laboratory Tests 1-15 

1.5 Standards Available for In Situ Testing 1-15 

1.6 Standards Available for Laboratory Testing of Soils 1-16 



1.1 Alignment of Boreholes 1-8 

1.2 Necessary Borehole Depths for Foundations 1-10 

1.3 Required Depth of Exploration 1-12 

1.4 Light Percussion Drilling Rig (Courtesy Of Pilcon Engineering Ltd) 1-18 

1.5 Light Percussion Drilling Tools 1-19 

1.6 Bucket Auger 1-19 

1.7 Selection of Hand-Operated Augers 1-20 

1.8 Washboring Rig (Based On Hvorslev 1949) 1-21 

1.9 Bits for Rotary Open Holing 1-22 

1.10 Sample Borelog indicating Logging of Soil and Rock in a Borehole 1-23 

1.11 Mackintosh Probe 1-25 

1-ii March 2009


1 PLANNING AND SCOPE 

1.1 INTRODUCTION 

One of the more important tasks to be considered, prior to carrying out soil investigations (SI) is to 

first understand clearly what is intended for the project in terms of design and construction, and the 

existing conditions of the site on which the project is to be established. Accordingly, where available, 

the requisite information to be had at the early stages of SI planning includes the detailed collection, 

inspection and study of the following: 

i. Topographic Maps: assist in or complement the examination of earthworks, soft ground and 

or or slope for site reconnaissance and planning of SI; 

ii. 

iii. 

iv. 

Geological Maps and Memoirs: assist with the planning of SI; methods of SI; and in deciding 

the extent of field and laboratory testing required or necessary; 

Site Histories: a good understanding and appreciation of the existence of old foundations, 

tunnel, underground services and etc. will provide for better SI planning; 

Results of Adjacent and Nearby SI: provide for a more efficient and economical SI; 

v. Details of Adjacent Structures and Foundations: provide for better safety assessment and 

prevention of foundation failure or settlement of adjacent properties due to current or 

proposed foundation works; and 

vi. 

Aerial Photographs: provide indication of geomorphological features, land use, problem areas 

and layout arrangements, and are particularly useful for highways and hillslope 

developments. 

1.2 GENERAL 

By general convention, site investigation can be defined as the process by which geological, 

geotechnical, and other relevant information which might affect the construction and performance of 

a civil engineering project is acquired. 

Due to the irregular nature of its deposition and its creation through the many processes out of a 

wide variety of materials, soils and rocks are notoriously variable, and often have properties which 

are undesirable from the point of view of a proposed structure. Often, the decision to develop a 

particular site cannot often be made on the basis of its complete suitability from the engineering 

viewpoint. Thus geotechnical problems may occur and require geotechnical parameters for their 

solution. 

1.3 OBJECTIVES 

Referring to the definitions as specified by the various Codes of Practices (BS CP 2001:1950, 1957; 

BS 5930:1981 & MS 2038:2006), the objectives of site investigation can be summarized and adopted 

herein as providing data for the following. 

i. Site selection. The construction of certain major projects, such as dams, is dependent on the 

availability of a suitable site. Clearly, if the plan is to build on the cheapest, most readily 

available land, geotechnical problems due to the high permeability of the sub-soil, or to slope 

instability may make the final cost of the construction prohibitive. Since the safety of lives and 

property are at stake, it is important to consider the geotechnical merits or demerits of 

various sites before the site is chosen for a project of such magnitude. 

March 2009 1-1


ii. 

iii. 

iv. 

Foundation and earthworks design. Generally, factors such as the availability of land at the 

right price, in a good location from the point of view of the eventual user, and with the 

planning consent for its proposed use are of over-riding importance. For medium-sized 

engineering works, such as expressways or highways and or or multi-storey structures, the 

geotechnical problems must be solved once the site is available, in order to allow a safe and 

economical design to be prepared. 

Temporary works design. The actual process of construction may often impose greater stress 

on the ground than the final structure. While excavating for foundations, steep side slopes 

may be used, and the in-flow of groundwater may cause severe problems and even collapse. 

These temporary difficulties, which may in extreme circumstances prevent the completion of 

a construction project, will not usually affect the design of the finished works. They must, 

however, be the object of serious investigation. 

The effects of the proposed project on its environment. The construction of an excavation 

may cause structural distress to neighbouring structures for a variety of reasons such as loss 

of ground, and lowering of the groundwater table. This will result in prompt legal action. On a 

wider scale, the extraction of water from the ground for drinking may cause pollution of the 

aquifer in coastal regions due to saline intrusion, and the construction of a major earth dam 

and lake may not only destroy agricultural land and game, but may introduce new diseases 

into large populations. These effects must be the subject of investigation. 

v. Investigation of existing construction. The observation and recording of the conditions leading 

to failure of soils or structures are of primary importance to the advance of soil mechanics, 

but the investigation of existing works can also be particularly valuable for obtaining data for 

use in proposed works on similar soil conditions. The rate of settlement, the necessity for 

special types of structural solution, and the bulk strength of the sub-soil may all be obtained 

with more certainty from back-analysis of the records of existing works than from small scale 

laboratory tests. 

vi. 

vii. 

The design of remedial works. If structures are seen to have failed, or to be about to fail, 

then remedial measures must be designed. Site investigation methods must be used to obtain 

parameters for design. 

Safety checks. Major civil engineering works, such as earth dams, have been constructed over 

a sufficiently long period for the precise construction method and the present stability of early 

examples to be in doubt. Site investigations are used to provide data to allow their continued 

use. 

By stipulation of the BS 5930: 1981 (and MS 2038:2006), site investigation aims to determine all the 

information relevant to site usage, including meteorological, hydrological and environmental 

information. Ground investigation on the other hand, aims only to determine the ground and 

groundwater conditions at and around the site through boring and drilling exploratory holes, and 

carrying out soil and rock testing. By common engineering convention, however, the terms site 

investigation and ground investigation can be used interchangeably. 

1.4 PHASES OF INVESTIGATIONS 

Site investigation work normally falls into three phases; i.e., reconnaissance, desk study and ground 

investigation, although these phases may be overlapped, merged or omitted, depending on site 

conditions and the requirements of a particular project. 

i. Reconnaissance: Involves visiting the site and its surroundings, and noting the salient 

features of the area; 

1-2 March 2009


ii. 

iii. 

Desk study: Includes a review of available information from aerial photographs, maps and 

records; and 

Ground investigation: Includes sinking pits and borings, field tests and observations, and 

laboratory testing. Geophysical surveys may also be helpful. 

As work proceeds, at any stage, the program may need to be modified in the light of the information 

obtained. The work involved in each of these stages of the site investigation procedures is discussed 

more fully in the following sections. 

1.5 APPROACHES TO SITE INVESTIGATIONS 

1.5.1 Approach 1: Reconnaissance – Site Visit 

Much useful information can be obtained simply by visiting the site and noting such features as 

topography, drainage, soil types, rock outcrops, vegetation, land use and the condition of existing 

roads, buildings and other structures. Details of former use of the site and nearby structures or 

proposed developments may also affect, or be affected by, the project, and should be considered. 

Examination of local quarries and cuttings and the limited use of geophysical techniques may also be 

appropriate. 

Site reconnaissance is necessary for the acquisition of the following (additional) information. 

i. To confirm and obtain additional information of the site; 

ii. 

iii. 

iv. 

To examine adjacent and nearby development: to record if any, the existence of predilapidation 

surveys, exposed cut slopes, appearance of cracks and settlements of adjacent 

buildings, etc., as with the case of the Batu Pond flood mitigation project; 

To compare the surface features and topography with data obtainable in the desk study, so 

that the presence of (any) cut and fill areas, as well as exposed services markings can be 

checked; 

To locate and study (any) outcrops and or or previous slips so that the corresponding 

inherent stability characteristics can be studied. 

1.5.2 Approach 2: Desk-Study and Geotechnical Advice 

The minimum requirement for a satisfactory investigation is that a desk study and walk-over survey 

are carried out by a competent geotechnical specialist, who has been carefully briefed by the lead 

technical construction professional (architect, engineer or quantity surveyor) as to the forms and 

locations of construction anticipated at the site. 

This approach will be satisfactory where routine construction (small scale construction which is not 

subjected to excessive loading of any kind, does not require elaborate and detailed designs and 

supervision) is being carried out in well-known and relatively uniform ground conditions. The desk 

study and walk-over survey are intended to: 

i. Confirm the presence of the anticipated ground conditions, as a result of the examination of 

geological maps and previous ground investigation records; 

ii. 

iii. 

Establish that the variability of the sub-soil is likely to be small; 

Identify potential construction problems; 

March 2009 1-3


iv. 

Establish the geotechnical limit states (for example, slope instability, excessive foundation 

settlement) which must be designed for; and to 

v. Investigate the likelihood of unexpected hazards (for example, made ground, or contaminated 

land). 

In this regard, it is unlikely that detailed geotechnical design parameters will be required, since the 

performance of the proposed development can be judged on the basis of previous construction. 

1.5.3 Approach 3: Ground Investigation 

Pits and Borings 

The choice of methods will depend on the depth to be investigated, the type of sampling required, 

the strata likely to be encountered and the resources available. The most common types of 

exploratory hole used in site investigation work are presented and described in subsequent chapters, 

along with illustrations of some types of drilling equipment in common use. 

Sampling 

Soil samples can generally be divided into two main categories; (i) disturbed samples and (ii) 

undisturbed samples. 

Disturbed samples include spoil from trial pit excavations, auger parings, sludge from a shell or from 

wash water return. The soil structure is disturbed and samples can be used only for classification 

tests or to determine the properties of remoulded soil. Small samples (500g) are usually put in jars 

or small polythene bags. Large samples (5-50 kg) are put in large, heavy duty polythene bags. 

Undisturbed samples contain blocks of soil which have been recovered in a more-or-less undisturbed 

state, retaining the natural soil structure and moisture content, although some sample disturbance is 

inevitable. In trial pits, blocks may be cut by hand but in boreholes special sampling devices are 

needed. 

A variety of sampling devices are available, aimed at recovering undisturbed samples in various 

subsoil conditions. The simplest is the open-ended sampler, used with shell and auger boring, for use 

in most c1ays. The main drawbacks of this sampler are that it is difficult to obtain samples in soft or 

very sandy clays; it does produce noticeable disturbance so that it is unsuitable for sampling soft or 

sensitive clays; and it is open to abuse by drillers who sometimes overdrive it in an attempt to obtain 

a full sample. Nevertheless, it is still by far the most common form of sampler for use in clays. 

In order to overcome the problems of recovery and sample disturbance in soft clays and clayey silts 

and sands, piston samplers are used. (The principles of tube and piston samplers are covered in later 

sections of this manual). 

Many other types and variations of sampling device have been developed, usually with the aims of 

reducing sample disturbance and recovering soft or sandy soils. However, sophisticated samplers are 

expensive and difficult to use and some sample disturbance is inevitable in boring and sampling 

operations. Because of these problems, in-situ tests are usually used in sands and soft clays. 

Probes 

Probes measure the resistance of the ground to a rod or cone which is forced into the soil. By far the 

most common probe is the standard penetration test (usua1ly abbreviated to SPT), in which a 

standard sample tube is driven into the soil by repeated blows of a standard falling hammer, or 

1-4 March 2009


monkey. The test is carried out in conjunction with shell and auger boring and rotary drilling. 

(Principal features of the equipment are given in subsequent chapters of this manual, along with 

notes on its use. Interpretation of the test is empirical and common correlations used to interpret 

test results are covered in subsequent chapter). 

Most other types of probes are used to penetrate the soil without the need for a borehole. Probes fall 

into two main categories: 

a. Dynamic cones, in which the probe is driven into the soil by means of a falling hammer. (Thus 

the SPT is a form of dynamic probing). For deeper penetration, without the use of a borehole, 

it is necessary to reduce skin friction between the soil and the rod being driven into the 

ground. Various methods are used to overcome the problem of skin friction. 

b. Static cones, which are jacked into the ground at a steady rate. Cone resistance and skin 

friction are measured separately, usually by providing a separate sleeve and incorporating 

strain gauges into the sleeve and tip. The results obtained can be correlated with bearing 

capacity and settlement factors for foundations. 

A small hand probe, known as the Mackintosh probe, consists simply of a standard probe head and 

connecting rods. The resistance of the soil is measured by counting the number of blows of a 

standard drop hammer which is required to drive it to a set distance (usually l50mm). The device is 

useful in that it gives a rough indication of subsoil conditions quickly, usually during preliminary 

exploration. 

1.6 EXPLORATION AND SAMPLING 

The site investigations should be carried out in a scientific, orderly and cost effective manner to 

determine the actual ground conditions at the site and to obtain the design parameters for 

engineering analysis and design. 

Because the planning of ground investigation is so important, it is essential that an experienced 

geotechnical specialist is consulted by the initiator of the project and his leading technical designer 

very early during conceptual design. 

Planning of a ground investigation can be broken down into its component parts as summarised in 

Table 1.1. 

The most important step in the entire process of site investigation is the appointment of a 

geotechnical specialist, at the early planning stage of a construction project. At present, much site 

investigation drilling and testing is carried out in a routine way, and in the absence of any significant 

plan. This can result in a significant waste of money, and time, since the work is carried out without 

reference to the special needs of the project. 

March 2009 1-5


Table 1.1 Planning a Ground Investigation 

Stage Action Responsibility of 

I Obtain the services of an experienced geotechnical Developer or client 

specialist 

II Carry out desk study and air photograph or LIDAR (if Geotechnical specialist 

available) interpretation to determine the probable 

ground conditions at the site 

III Conceptual design: optimize construction to minimize 

geotechnical risk 

Architect, structural engineer, 

geotechnical specialist 

IV Identify parameters required for detailed Geotechnical specialist 

geotechnical calculations 

V Plan ground investigation to determine ground Geotechnical specialist 

conditions, and their variation, and to obtain 

geotechnical parameters. 

VI Define methods of investigation and testing to be Geotechnical specialist 

used 

VII Determine minimum acceptable standards for Geotechnical specialist 

ground investigation work 

VIII Identify suitable methods of procurement Geotechnical specialist, lead 

professional 

design, developer or client 

1.6.1 Spacing of Pits and Borings 

The required spacing depends very much on the size and type of the project and on the terrain and 

subsurface conditions. For a start, borings should initially be widely spaced and subsequently, 

intermediate borings can be carried out as required, so that sections can be drawn with reasonable 

accuracy. In uniform conditions, spacing may be 25m to 150m or more but spacings of 10m or less 

may be required to examine detailed problems and or or in erratic conditions. Examples of typical 

spacing requirements are given in Table 1.2 and illustrated in Fig 1.1. Where structures are to be 

founded on slopes, the overall stability of the structure and the slope must obviously be investigated, 

and to this end a deep borehole near the top of the slope will be very useful. 

It must be emphasised however, that the requirements of individual sites may vary considerably 

from those given. 

1-6 March 2009


Table 1.2 Recommended Number and Depth of Borings 

LOCATION TO BE 

INVESTIGATED 

NEW SITE OF 

FAIRLY WIIDE 

EXTENT 

FOUNDATIONS 

FOR 

STRUCTURES 

Low-rise, 1 or 2 

Storey Buildings 

Multi-storey 

Buildings 

Buildings on Poor 

or Variable 

Grounds 

Bridge piers, 

Abutments, 

STABILITY 

SLOPES 

ROADS, 

RUNWAYS 

PIPELINES 

OF 

AND 

BORROW PITS 

(for compacted 

fill) 

DISTANCE BETWEEN BORINGS 

(m) 

Horizontal Stratification of Soil 

Uniform Average Erratic 

MINIMUM 

NUMBER OF 

BORINGS 

REQUIRED 

(nos.) 

RECOMMENDED 

MINIMUM DEPTH 

- - - 5 to 10 - 

60 30 15 

45 30 15 

- - - 

- 30 7.5 

- - - 

- - 

1 to 3 for 

each structure 

2 to 4 for 


2 to 4 for 


1 to 3 for 

each pier or 

abutment 

3 to 5 along 

each critical 

section 

250 150 30 - 

300 - 

150 

150 - 60 30 - 15 - - 

1.5 times width of 

loaded or plan area 

1.5 times width of 

loaded or plan area 

or up to 6m into 

firm or hard layer 

or 3m into 

bedrock, whichever 

encountered earlier 

Up to 9m into firm 

or hard layer or 

4.5m into bedrock, 

whichever 

encountered layer 

Up to 10.5m into 

firm or hard layer 

or 6m into 

bedrock, whichever 

encountered layer 

Below slip plane or 

6m into firm or 

hard layer or 3m 

into bedrock, 

whichever 

encountered earlier 

2m to 3m below 

formation for 

roads, 6m below 

formation for 

runways, 0.5m 

below invert for 

pipelines 

March 2009 1-7


140 

‘A’ 

BH1 

Site boundary 

130 

120 

BH2 

110 

BH6 BH3 BH7 

100 

90 

BH4 

Probable position 

of structure 

60 

BH5 

‘A’ 

(a) Site plan 

BH1 

BH2 

BH3 

BH5 

BH4 

(b) Section ‘A’ – ‘A’ 

Figure 1.1 Alignment of Boreholes 

1-8 March 2009


1.6.2 Depths of Borings 

The required depths depend mainly on the subsoil conditions and on the type of proposed structure 

or development. Where poor foundation material, such as soft clay, loose sand or uncompacted fill, 

is encountered, borings should be extended through this to reach sounder material. If great depths 

of soft, compressible or loose material are encountered, borings should be taken down to a depth 

where the imposed stress from the proposed structure is negligible. 

Where good conditions are encountered at shallow depths, borings should be taken to a depth 

where the possible presence of weaker material below the depth explored would not seriously affect 

the proposed structure. Where bedrock is encountered, borings should extend typically about l.5m 

into sound rock and 3-5m into weathered rock, though this. will depend on site conditions and will 

be inadequate, for instance, where old mine workings may be present. At least one boring should 

extend well below the zones normally investigated, as a check on the conditions at depth. 

As a rough guide to the necessary depths, as determined from considerations of stress distribution or 

seepage, the following depths may be used. 

1. Reservoirs. Explore soil to: (i) the depth of the base of the impermeable stratum, or (ii) not 

less than 2 x maximum hydraulic head expected. 

2. Foundations. Explore soil to the depth to which it will be significantly stressed. This is often 

taken as the depth at which the vertical total stress increase due to the foundation is equal to 

10% of the stress applied at foundation level (Fig. 1.2). 

3. For roads. Ground exploration need generally only proceed to 2 - 4 m below the finished road 

level, provided the vertical alignment is fixed. In practice some realignment often occurs in 

cuttings, and side drains may be dug up to 6 m deep. If site investigation is to allow flexibility 

in design, it is good practice to bore to at least 5 m below ground level where the finished 

road level is near existing ground level, 5 m below finished road level in cut, or at least oneand-a 

half times the embankment height in fill areas. 

4. For dams. For earth structures, Hvorslev (1949) recommends a depth equal to one-half of the 

base width of the dam. For concrete structures the depth of exploration should be between 

one-and-a-half and two times the height of the dam. Because the critical factor is safety 

against seepage and foundation failure, boreholes should penetrate not only soft or unstable 

materials, but also permeable materials to such a depth that seepage patterns can be 

predicted. 

5. For retaining walls. It has been suggested by Hvorslev that the preliminary depth of 

exploration should be three-quarters to one-and-a-half times the wall height below the bottom 

of the wall or its supporting piles. Because it is rare that more than one survey will be carried 

out for a small structure, it will generally be better to err on the safe side and bore to at least 

two times the probable wall height below the base of the wall. 

6. For embankments. The depth of exploration should be at least equal to the height of the 

embankment and should ideally penetrate all soft soils if stability is to be investigated. If 

settlements are critical then soil may be significantly stressed to depths below the bottom of 

the embankment equal to the embankment width. 

The general required depths of ground exploration for the various engineering structures are further 

illustrated in Fig. 1.3. 

March 2009 1-9


S 

S 

BH 

D 

B 

D 

B 

Borehole depth 

>[D+1.5x8] 


>[D+1.5(25+B)] 

For S < 5B 

a) Structure on isolated pad or raft 

b) Closely spaced strip on pad footings 

B 

Notional equivalent 

raft at 2/3 depth 

D 

Individual 

pressure bulbs 


>[2/3 D+1.58] 

Combined 

pressure bulb 

c) Large structure on friction piles 

Figure 1.2 Necessary Borehole Depths for Foundations 

1-10 March 2009


2L 

H 

Dams/Reservoirs/ 

Levees 

D 

D = Impermeable Stratum or Bedrock, or Not less than 2 x maximum hydraulic head expected, or ½ 

H- 2H 

B 

B 

L 

Unit load 

P 

Total load 

P=P.L.B. 

L 

L1 

B1 

P1 

S1 

S1 

Foundation 

Structure 

D 

MAT OR SINGLE FOOTING 

S 

S 

GROUP OF FOOTINGS 

D = 2B (square) to 6B (strip) 

D 

D 

Roads/ 

Farm Roads 

(i) Roads: At least 5m below finished road level (near existing ground and in cut 

(ii) Farm Roads: D = 1m to 2m (light traffic); 2m to 3m (heavy traffic) 

Figure 1.3 (a) 

March 2009 1-11


H 

Retaining & 

Quay Walls 

D 

D = 2H to 3H 

L 

H 

2L 

D 

Terraces/Fill 

Embankments 

H 

D 

D = 2L (embankment) to 4L (terraces) 

H 

Deep Cuts 

D = 2B to 4B 

Figure 1.3 (b) 

Figure 1.3 Required Depth of Exploration 

Because many investigations are carried out to determine the type of foundations that must be used, 

all borings should be carried to a suitable bearing strata, and a reasonable proportion of the holes 

should be planned on the assumption that piling will have to be used. 

1.6.3 Sampling, Laboratory Testing and In situ Testing Requirements 

D 

The types and spacing of samples depends on the material encountered and the type of project 

undertaken. As a general guide, undisturbed samples in clays or standard penetration tests in sands 

should be carried out at l.5m to 3m intervals and at every change in stratum, in shell and auger 

borings. Standard or cone penetration tests should be carried out every l.5m in rotary drillholes 

B 

1-12 March 2009


through sand and gravel. Disturbed samples however, should be taken in all kinds of borings at 1.5m 

intervals and at each change of stratum. 

Accordingly, the sampling routine should be aimed at: 

i. Providing sufficient samples to classify the soil into broad soil groups, on the basis of particle 

size and compressibility; 

ii. Assessing the variability of the soil; 

iii. Providing soil specimens of suitable quality for strength and compressibility testing; and 

iv. Providing specimens of soil and groundwater for chemical testing. 

In soft clays or for special conditions, continuous sampling may be necessary. Excessive use of water 

to advance borings in clays should be avoided and, before a sample is taken, the bottom of the 

borehole should be carefully cleaned out. 

Undisturbed samples should be kept sealed with wax. Bulk samples are usually stored in heavy-duty 

polythene bags tied up tightly with string. Small disturbed samples, usually taken from the cutting 

shoe of an open-ended sampler or from the split-spoon sampler used in the standard penetration 

test, are kept in jars, tins or small polythene bags. Water samples should be taken whenever water 

is encountered during drilling. Samples are stored in jars whose lids are sealed by dipping them in 

paraffin wax. 

All samples must be clearly labelled, with labels both inside and outside the containers, and must be 

carefully transported and stored. Once they are no longer required for inspection or testing, samples 

may be discarded. However, care should be taken that they are not discarded too soon and all the 

people who may wish to make use of the samples should be informed before they are disposed of. 

In situ testing is carried out when: 

i. Good quality sampling is impossible (for example, in granular soils, in fractured rock masses, in 

very soft or sensitive clays, or in stoney soils); 

ii. The parameter required cannot be obtained from laboratory tests (for example, in situ 

horizontal stress); 

iii. When in situ tests are cheap and quick, relative to the process of sampling and laboratory 

testing (for example, the use of the spt in clay, to determine undrained shear strength); and 

most importantly, 

iv. For profiling and classification of soils (for example, with the cone test, or with dynamic 

penetration tests). 

The most commonly used test is the Standard Penetration Test (SPT), which is routinely used at 1.5 

m intervals within boreholes in granular soils, stoney soils, and weak rock. Other common in situ 

tests include the field vane (used only in soft and very soft cohesive soils), the plate test (used in 

granular soils and fractured weak rocks), and permeability tests (used in most ground, to determine 

the coefficient of permeability). 

The primary decision will be whether to test in the laboratory or in situ. Table 1.3 gives the relative 

merits of these options. 

March 2009 1-13


Table 1.3 Relative Merits of In Situ and Laboratory Testing 

In situ testing 

Test results can be obtained during the 

course of the investigation, much earlier 

than laboratory test results 

Appropriate methods may be able to test 

large volumes of ground, ensuring that the 

effects of large particle sizes and 

discontinuities are fully represented 

Estimates of in situ horizontal stress can be 

obtained 

Drainage boundaries are not controlled, so 

that it cannot definitely be known whether 

loading tests are fully undrained 

Stress path and or or strain levels are often 

poorly controlled 

Tests to determine effective stress strength 

parameters cannot be made, because of the 

expense and inconvenience of a long test 

period 

Pore pressures cannot be measured in the 

tested volume, so that effective stresses are 

unknown. 

Advantages 

Disadvantages 

Laboratory testing 

Tests are carried out in a well-regulated 

environment 

Stress and strain levels are controlled, as 

are drainage boundaries and strain rates 

Effective strength testing is straightforward 

The effect of stress path and history can be 

examined 

Drained bulk modulus can be determined 

Testing cannot be used whenever samples 

of sufficient quality and size are obtainable, 

for example, in granular soils, fractured 

weak rock, stoney clays 

Test results are only available some time 

after the completion of fieldwork 

The ground investigation planner requires a detailed and up-to-date knowledge of both laboratory 

and in situ testing, if the best choices are to be made. Table 1.4 gives a summary of the local current 

situation — but this will rapidly become out of date. Whatever is used depends upon the soil and 

rock encountered, upon the need (profiling, classification, parameter determination), and upon the 

sophistication of geotechnical design that is anticipated. 

1-14 March 2009


Table 1.4 Common Uses of In Situ and Laboratory Tests 

Purpose Suitable laboratory test Suitable in situ test 

Profiling 


Particle size distribution 

Plasticity (Atterberg limits) 

Undrained strength 

Cone test 

Dynamic penetration test 

Geophysical down-hole 

logging 

Classification 


Plasticity (Atterberg limits) 

Cone 

Parameter 

determination: 

Undrained strength, 

cu 

Peak effective 

strength, c’ φ’ 

Residual strength, 

c’ φ’ 

Compressibility 

Permeability 

Chemical 

characteristics 

Undrained triaxial 

Effective strength triaxial 

Shear box 

Ring shear 

Oedometer 

Triaxial, with small strain 

measurement 

Triaxial consolidation 

Triaxial permeability 

pH 

Sulphate content 

SPT 

Cone 

Vane 

Self-boring pressuremeter 

Plate test 

In situ permeability tests 

Geophysical resistivity 

The following table (Table 1.5 refers) details the applicable standards available for in-situ testing, 

while Table 1.6 details on standards available for laboratory soils testing. 

Table 1.5 Standards Available for In Situ Testing 

Test British Standard American Standard 

Density tests (sand BS 1377: part 9: 1990, clause 2 ASTM D1556-82 

replacement, 

water 

ASTM D2937-83 

replacement, 

core 

ASTM D2937-84 

cutter,balloon and nuclear 

ASTM D2922-91 

methods) 

Apparent resistivity BS 1377: part 9: 1990, clause 5.1 ASTM G57-78 (reapproved 

1984) 

In situ redox potential BS 1377: part 9: 1990, clause 5.2 

In situ California bearing ratio BS 1377: part 9: 1990, clause 4.3 ASTM D4429-84 

Standard penetration test BS 1377: part 9: 1990, clause 3.3 ASTM D1586-84 

ASTM D4633-86 (energy 

measurement) 

Dynamic penetration test BS 1377: part 9: 1990, clause 3.2 

Cone penetration test BS 1377: part 9: 1990, clause 3.1 ASTM D3441-86 

Vane test BS 1377: part 9: 1990, clause 4.4 ASTM D2573-72 

(reapproved 1978) 

Plate loading tests 

BS 1377: part 9: 1990, clause ASTM D1194-72 

4.1, 4.2 


ASTM D4395-84 

Pressuremeter test ASTM D4719-87 

March 2009 1-15


Atterberg limits 

Density 

Specific gravity 


Pinhole dispersion test 

Table 1.6 Standards Available for Laboratory Testing of Soils 

Test British Standard American Standard 

Classification tests 


BS 1377:part 2:1990, clause 3 ASTM D2216-91 

ASTM D4643-87 

ASTM D4318-84 

Organic matter content 

Loss on ignition 

Sulphate content 

Carbonate content 

Chloride content 

pH 

Resistivity 

Redox potential 

Proctor or 2.5kg rammer 

Heavy or 4.5kg rammer 

Vibrating hammer 

California bearing ratio 

Undrained triaxial shear 

strength 

Effective strength from the 

consolidated-undrained triaxial 

compression test with pore 

pressure measurement 

Effective strength from the 

consolidated-drained triaxial 

compression test with volume 

change measurement 

Residual strength by direct 

shear testing in the shear box 

Residual strength using the 

Bromhead ring shear apparatus 

One-dimensional compressibility 

in the oedometer 

Isotropic consolidation in the 

triaxial apparatus 

BS 1377:part 2:1990, clause 4, 5 

BS 1377:part 2:1990, clause 7 



Chemical tests 







BS 1377:part 3:1990, clause 10 

BS 1377:part 3:1990, clause 11 

Compaction tests 

BS 1377:part 4:1990, clause 3.3 

BS 1377:part 4:1990, clause 3.5 

BS 1377:part 4:1990, clause 3.7 

Strength tests 


BS 1377:part 7:1990, clause 8, 9 




Compressibility tests 

BS 1377:part 5:1990, clause 3, 4 


ASTM D854-92 

ASTM D422-63 (reapproved 

1972) 

ASTM D2217-85 

ASTM D4647-87 

ASTM D2974-87 

ASTM D4373-84 

ASTM G51-77(reapproved 

1984) 

ASTM D698-91 

ASTM D1557-91 

ASTM D1883-92 

ASTM D2850-87 

ASTM D3080-90 

ASTM D2435-90 

Permeability tests 

In the constant-head apparatus BS 1377:part 5:1990, clause 5 ASTM D2434-68 


1-16 March 2009


The key points in checking the effectiveness of a site investigation are as follows. 

1. Avoid excessive disturbance. Look for damaged cutting shoes, rusty, rough or dirty sample 

barrels, or badly designed samplers. Check the depth of casings to ensure that these never 

penetrate beneath the bottom of the borehole. Try to assess the amount of displacement 

occurring beneath power augers, and prevent their use if necessary. 

2. Check for water. Ensure that adequate water levels are maintained when drilling in granular 

soils or soft alluvium beneath the water table. The addition of water in small quantities should 

be kept to a minimum, since this allows swelling without going any way towards replacing total 

stress levels. Make sure the driller stops drilling when groundwater is met. 

3. Check depths. The depths of samples can be found approximately by noting the number of rods 

placed on the sampling tool as it is lowered down the hole, and the amount of ‘stick-up’ of the 

last rod at the top of the hole. This type of approach is often used by drillers, but is not always 

satisfactory. Immediately before any sample is taken or in situ test performed the depth of the 

bottom of the hole should be measured, using a weighted tape. If this depth is different from 

the last depth of the drilling tools then either the sides of the hole are collapsing, or soil is 

piping or heaving into the base. Open-drive sampling should not then be used. 

4. Look for faulty equipment. On-site maintenance may lead to SPT hammers becoming nonstandard, 

for example owing to threading snapping and the central stem being shortened, 

giving a short drop. When working overseas with subcontract rigs the weight of the SPT 

hammer should also be measured. Other problems which often occur are: (i) the blocking of 

vents in sampler heads; and (ii) the jamming of inner barrels in double tube swivel-type 

corebarrels. 

5. Examine driller’s records regularly. The driller should be aware that the engineer is seeking high 

quality workmanship. One of the easiest ways of improving site investigation is to demand that 

up to the moment records are kept by the driller as drilling proceeds. These should then be 

checked several times a day when the engineer visits the borehole. Any problems encountered 

by the driller can then be discussed, and decisions taken. 

1.7 METHODS OF SITE INVESTIGATION – DRILLING AND SAMPLING 

The next phase of the SI planning involves an appreciable understanding of the different methods 

commonly available for the local SI practices, and their corresponding use and limitations. This 

chapter briefly describes the equipment and procedures commonly used for the drilling and sampling 

of soil and rock. The methods addressed in this chapter are used to retrieve soil samples and rock 

cores for visual examination and laboratory testing. 

1.7.1 Subsurface Exploration 

The primary functions of any ground investigation process will be one of the following: 

i. Locating specific ‘targets’, such as dissolution features or abandoned mineworkings 

ii. Determining the lateral variability of the ground; 

iii. Profiling, including the determination of groundwater conditions; 

iv. Index testing; 

v. Classification; 

vi. Parameter determination. 

March 2009 1-17


1.7.2 

Boring 

Numerous methods are available for advancing 

boreholes to 

obtain samples or details of soil strata. 

The particular methods used by any country will tend to be 

restricted, based on their suitability for 

local ground conditions. The principal methods used worldwide include: 

• Light 

percussion drilling; 

• Power augering; and 

• Washboring. 

1.7.2.1 

Light Percussion Drilling 

Often called ‘shell and auger’ drilling, this method is more properly termed light percussion drilling 

since the 

barrel auger is now rarely used with this type of equipment. The drilling 

rig (Fig. 1. 4) 

consists of: 

i. A collapsible ‘A’ frame, with a pulley at its top; 

ii. A diesel engine; connected via 

a hand-operated friction clutch (based on a brake 

drum system) 

to 

iii. A winch drum which provides pulling power to the rig rope and can 

be held stilll with a friction 

brake 

which is foot-operated. 

In clays, progress is made by dropping a steel tube knownn as a ‘claycutter’ into the 

soil (see Fig. 

1.5). This is slowly pulled out of the borehole and is then generally found 

to have soil wedged inside 

it. 

Figure 1. .4 Light Percussion Drilling 

Rig (Courtesy of Pilcon Engineering Ltd) 

1-18 

March 2009


Figure 1.5 Light Percussion Drilling Tools 

1.7.2.2 

Augering 

Augers may be classified as either bucket augers (Fig. 1.6) or flight augers. Bucket augers are similar 

in construction to the 

flat-bottomed Sprague and Henwood 

barrel auger. They consist of an open- 

break up 

the soil and allow it to enter the bucket as it is rotated. The top 

of the bucket is connected 

to a rod which transmits the torque and downward pressuree from the rig 

at ground level to the base 

of the hole: this rod is 

termed a ‘Kelly’. 

topped cylinder whichh has a base 

plate with one or two slots reinforced with cutting teeth, which 

Figure 1. 6 Bucket Auger 

The hand auger provides a light, portable method of sampling soft to 

surface. At least six types of auger 

are readily available: 

stiff soils near the ground 

• Posthole or Iwan auger; 

• Small helical auger (wood auger); 

• Dutch auger; 

• Gravel auger; 

March 2009 

1-19


• Barrel auger; and 

• Spiral auger. 

Figure 1.7 shows a selection of these augers. 

1.7.2.3 Wash Boring 

Figure 1.7 Selection of Hand-Operated Augers 

Wash boring is a relatively old method of boring small-diameter exploratory holes in fine-grained 

cohesive and non-cohesive soils. It was widely used in the USA in the first half of this century, but 

has been largely replaced by power auger methods. It is still used in areas of the world where labour 

is relatively cheap, for example southern Brazil. 

A very light tripod is erected, and a sheave is hung from it (Fig. 1.8). In its simplest form there are 

no motorized winches and the drilling water is pumped either by hand, or by a small petrol-driven 

water pump. Hollow drilling rods are connected to the pump via a flexible hose, and the drilling crew 

lift the string of rods by hand, or using a ‘cathead’ (a continuously rotating steel drum, around which 

a manilla rope is wound). 

1-20 March 2009


Figure 1.8 Washboring Rig (Based On Hvorslev 1949) 

1.7.3 Drilling 

Rotary drilling uses a rotary action combined with downward force to grind away the material in 

which a hole is being made. Rotary methods may be applied to soil or rock, but are generally easier 

to use in strong intact rock than in the weak weathered rocks and soils that are typically encountered 

during ground investigations. For a detailed description of equipment and methods the reader is 

referred to Heinz (1989). 

1.7.3.1 Open-Holing 

Rotary methods may be used to produce a hole in rock, or they may be used to obtain samples of 

the rock while the hole is being advanced. The formation of a hole in the subsoil without taking 

intact samples is known as ‘open-holing’. It can be carried out in a number of ways, but in site 

investigation a commonly used tool is the ‘tricone rock roller bit’ (or roller core bit) (Fig. 1.9). 

1.7.3.2 Coring 

The most common use of rotary coring in ground investigations is to obtain intact samples of the 

rock being drilled, at the same time as advancing the borehole. To do this a corebarrel, fitted with a 

‘corebit’ at its lower end, is rotated and grinds away an annulus of rock. The stick of rock, the ‘core’, 

in the centre of the annulus passes up into the corebarrel, and is subsequently removed from the 

borehole when the corebarrel is full. The length of core drilled before it becomes necessary to 

remove and empty the corebarrel is termed a ‘run’. 

March 2009 1-21


Figure 1.9 Bits for Rotary Open Holing 

Figure 1.10 shows the logging of soil and rock with in a borelog. 

1-22 March 2009


KKK 

BBB 

Figure 1.10 Sample Borelog indicating Logging of Soil and Rock in a Borehole 

March 2009 

1-23


1.7.4 Exploration Pit Excavation 

Exploration pits and trenches permit detailed examination of the soil and rock conditions at shallow 

depths and relatively low cost. Exploration pits can be an important part of geotechnical explorations 

where significant variations in soil conditions occur (vertically and horizontally), large soil and or or 

non-soil materials exist (boulders, cobbles, debris) that cannot be sampled with conventional 

methods, or buried features must be identified and or or measured. 

Exploration pits are generally excavated with mechanical equipment (backhoe, bulldozer) rather than 

by hand excavation. The depth of the exploration pit is determined by the exploration requirements, 

but is typically about 2 m (6.5 ft) to 3 m (10 ft). In areas with high groundwater level, the depth of 

the pit may be limited by the water table. Exploration pit excavations are generally unsafe and or or 

uneconomical at depths greater than about 5 m (16 ft) depending on the soil conditions. 

1.7.5 Probing 

A wide range of dynamic and static penetrometers are available, with different types being used in 

different countries. However, the objective of all probing is the same, namely to provide a profile of 

penetration resistance with depth, in order to give an assessment of the variability of a site. Probing 

is carried out rapidly, with simple equipment. It produces simple results, in terms of blows per unit 

depth of penetration, which are generally plotted as blowcount or depth graphs 

1.7.5.1 MacKintosh Probe 

The Mackintosh prospecting tool (also commonly known as JKR probe) consists of rods which can be 

threaded together with barrel connectors and which are normally fitted with a driving point at their 

base, and a light hand-operated driving hammer at their top (Fig. 1.10). The tool provides a very 

economical method of determining the thickness of soft deposits such as peat. 

The driving point is streamlined in longitudinal section with a maximum diameter of 27mm. The drive 

hammer has a total weight of about 5kg. The rods are 1.2 m long and 12mm dia. The device is often 

used to provide a depth profile by driving the point and rods into the ground with equal blows of the 

full drop height available from the hammer: the number of blows for each 300 mm of penetration is 

recorded. When small pockets of stiff clay are to be penetrated, an auger or a core tube can be 

substituted for the driving point. The rods can be rotated clockwise at ground level by using a box 

spanner and tommy bar. Tools can be pushed into or pulled out of the soil using a lifting or driving 

tool. Because of the light hammer weight the Mackintosh probe is limited in the depths and materials 

it can penetrate. 

In Malaysia, this method of investigation is usually employed during preliminary investigative works. 

It involves the use of: 

• 5 kg hammer weight, 

• Dropped from a guided free fall height of 280mm (JKR probe), and 

• Usually carried out up to a depth of 12m, or upon encountering the 400 resistance blows or 300 

mm. 

The test itself is relatively cheap and quick to execute, and is used to establish: 

• Localised soft area or weak layer or spot or slip plane; 

• The presence of hard or bearing layers or shallow bedrocks, as in the case of limestone profiling; 

• Preliminary subsoil information (eg. soil consistency & undrained shear strength, c u ); and 

• The interpolation between boreholes or piezocones. 

1-24 March 2009


Limitations associated with this test include: 

• Relatively shallow test depths (deeper depths in coarse materials give misleading results); and 

• Prone to human errors: variation in drop weight or exerting force, gives rise to misleading 

results, and risks of wrong counting unless mechanical counter is used. 

Precautionary measures to be observed require that: 

• The drop of the hammer should be free falling and consistent with each drop height; and 

• The components and apparatus must be properly washed and oiled. 

1.7.6 Examination In-Situ 

1.7.6.1 Trial Pit 

Figure 1.11 Mackintosh Probe 

Trial pits provide the best method of obtaining very detailed information on strength, stratification, 

pre-existing shear surfaces, and discontinuities in soil. Very high quality block samples can be taken 

only from trial pits. 

It is as well to note that every year many people are killed during the collapse of unsupported 

trenches. Remember to be careful — do not enter trenches or pits more than 1.2m deep without 

either supporting the sides or battering back the sides. Even so, if a pit is dug and remains stable 

without support, a quick means of exit such as a ladder should be provided. 

Trial pits may be excavated by either hand digging or machine excavation. In general, machine 

excavation is used for shallow pits, whereas hand excavation is used for deep pits which must be 

supported. In the limited space of a trial pit, which is often 1.5m x 3m in plan area at ground level, it 

is usually impossible to place supports as machine excavation proceeds. Shallow trial pits provide a 

March 2009 1-25


cheap method of examining near-surface deposits in situ, but the cost increases dramatically with 

depth, because of the need to support. 

Shallow trial pits can be excavated by wheeled offset backhoe which has a digging depth of about 

3.5 – 4.0m, and may not be able to move easily across wet steeply sloping sites. Deeper pits, or pits 

where access is difficult can be excavated by 360° slew-tracked hydraulic excavators. These 

machines have a digging depth of about 6 m, and an available digging force about 50—100% 

greater than the backhoe. 

1-26 March 2009


REFERENCES 

[1] Acker, W. L., III (1974). Basic Procedures for Soil Sampling and Core Drilling, Acker Drill Co. 

Inc., P.O. Box 830, Scranton, PA., 18501. 

[2] ADSC (1995). “Recommended procedures for the entry of drilled shaft foundations 

excavations.” The International Association of Foundation Drilling, (IAFD-ADSC), Dallas. 

[3] Contract DACW39-86-M-4273, Department of the Army, U.S. Army Corps of Engineers, 

Washington, D.C. 

[4] Hunt, R. E. (1984). Geotechnical Engineering Investigation Manual, McGraw-Hill Inc., 983 p. 

[5] Leroueil, S. and Jamiolkowski, M. (1991). “Exploration of soft soil and determination of 

design parameters”, Proceedings, GeoCoast’91, Vol. 2, Port & Harbor Res. Inst., Yokohama, 969-998. 

[6] Lowe III, J., and Zaccheo, P.F. (1991). "Subsurface explorations and sampling." Foundation 

Engineering Handbook, H. Y. Fang, ed., Van Nostrand Reinhold, New York, 1-71. 

[7] Lutenegger, A. J., DeGroot, D. J., Mirza, C., and Bozozuk, M. (1995). “Recommended 

guidelines for sealing geotechnical exploratory holes.” FHWA Report 378, Federal Highway 

Administration Washington, D.C. 

[8] Skempton, A. W. (1957). Discussion on “The planning and design of new Hong Kong 

airport.” Proceedings, Institution of Civil Engineers, Vol. 7 (3), London, 305-307. 

[9] U.S. Department of the Interior, Bureau of Reclamation. (1973). Design of small dams, 

United States Government Printing Office, Washington, D.C. 

[10] U.S. Department of the Interior, Bureau of Reclamation (1960). Earth manual, United States 

Government Printing Office, Washington, D.C. 

March 2009 1-27



1-28 March 2009

CHAPTER 2 SAMPLING AND SAMPLING DISTURBANCE

Chapter 2 SAMPLING AND SAMPLING DISTURBANCE 





2.1 INTRODUCTION .......................................................................................................... 2-1 

2.2 SAMPLING METHODS ................................................................................................... 2-1 

2.2.1 Undistured Sample ........................................................................................ 2-1 

2.2.2 Disturbed Sampling ....................................................................................... 2-4 

2.3 SAMPLING INTERVAL AND APPROPRIATE SAMPLER TYPE ............................................... 2-5 

2.4 SAMPLE RECOVERY ..................................................................................................... 2-5 

2.5 REQUIRED VOLUME OF MATERIAL FOR TESTING PROGRAMME ...................................... 2-5 

2.6 SAMPLE DISTURBANCE ................................................................................................ 2-7 

REFERENCES ....................................................................................................................... 2-10 

March 2009 2-i


List of Table 


2.1 Common Sampling Methods 2-2 

2.2 Mass of Disturbed Soil Sample Required For Various Tests 2-7 



2.1 Effects of Tube Sampling Disturbance of Lightly Overconsolidated Natural 

(‘Structured’) 2-8 

2.2 Influence of Tube Sampling Disturbance on Undrained Strength and Stiffness 2-9 

2-ii March 2009



2 SAMPLING AND SAMPLING DISTURBANCE 

Sampling is soil and rock is carried out for identification and description of soils strata and rock type 

with depth, and to perform laboratory testing for determination of index, classification and 

engineering properties. Laboratory tests typically consist of: 

i. Index tests (for example, unconfined compressive strength tests on rock); 

ii. Classification tests (for example, Atterberg limit tests on clays); and 

iii. Tests to determine engineering design parameters (for example strength, compressibility, 

and permeability). 

Samples obtained either for description or testing should be representative of the ground from which 

they are taken. They should be large enough to contain representative particle sizes, fabric, and 

fissuring and fracturing. They should be taken in such a way that they have not lost fractions of the 

in situ soil (for example, coarse or fine particles) and, where strength and compressibility tests are 

planned, they should be subject to as little disturbance as possible. 

2.2 SAMPLING METHODS 

Generally, sampling during a soil investigation program can be grouped into two main categories. 

1. Undisturbed sampling 

2. Disturbed sampling 

The methods of sampling adopted for a particular site investigation program is based on the type 

and requirement of soil investigation data for design and construction. While a large number of 

samplers and sampling methods are available, however, before a suitable technique can be selected 

it is always necessary to consider whether the sample size will be adequate, and whether the most 

suitable method of sampling has been selected, to ensure that sample disturbance is sufficiently 

small. 

2.2.1 Undistured Sample 

Undisturbed samples are obtained with specialized equipment designed to minimize the disturbance 

to the in-situ structure and moisture content of the soils. The term “undisturbed” soil sample refers 

to the relative degree of disturbance to the soil’s in-situ properties. Specimens obtained by 

undisturbed sampling methods are used to determine the strength, stratification, permeability, 

density, consolidation, dynamic properties, and other engineering characteristics of soils. 

Undisturbed samples are obtained in clay soil strata for use in laboratory testing to determine the 

engineering properties of those soils. Undisturbed samples of granular soils can be obtained, but 

often specialized procedures are required such as freezing or resin impregnation and block or core 

type sampling. Common methods for obtaining undisturbed samples are summarized in Table 2.1 

and briefly described below. 

March 2009 2-1


Table 2.1 Common Sampling Methods 

Sampler Disturbed/ 

Undisturbed 

Appropriate Soil Types Method of Penetration % Use in 

Practice 

Split-Barrel Disturbed Sands, silts, clays Hammer driven 85 

(Split Spoon) 

Thin-Walled Undisturbed Clays, silts, fine-grained Mechanically Pushed 6 

Shelby Tube 

soils, clayey sands 

Continuous Partially Sands, silts, and clays Hydraulic push with 4 

Push Undisturbed 

plastic lining 

Piston Undisturbed Silts and clays Hydraulic push 1 

Pitcher Undisturbed Stiff to hard clay, silt, 

sand, partially weathered 

rock and frozen or resin 

impregnated granular soil 

Rotation and hydraulic 

pressure 


Denison Sampler 

The Denison sampler was designed by H.L. Johnson in 1930 to obtain samples from dense or highly 

cemented strata (stiff to hard clays and dense sand) where thin Shelby tube was unable to penetrate 

and extract an undisturbed sample. The sample device is essentially a double tube core barrel with 

thin lined liner tube adapted to soil use. The inner tube with cutting shoe always advances ahead of 

the rotating outer core barrel ensuring the sample to e undisturbed and uncontaminated. 

The Denison core barrel is manufactured in 89, 100, 140, and 197mm OD sizes, and recovers 

relatively large samples in the inner stationary tube. The standard lengths are 60cm and 1.5m. 

Besides stiff and dense soils, the sampler can also sample clean sand and soft clays with use of 

drilling mud, vacuum valve and basket core retainer. The operating procedure is to lower the 

sampler to the bottom of the hole and apply hydraulic feed downward pressure, simultaneously 

drilling at a maximum rate of 100 rpm and allowing the circulation of drilling fluid just enough to 

wash the cutting. Once the depth is reached, the core barrel is withdrawn, the head and cutting shoe 

is removed and inner liner pushed hydraulically or mechanically from inner core barrel. The soil 

sample collected in the liner is sealed in th same was as Shelby tube and the sample is logged. 

Pitcher Sampler 

The pitcher sampler is basically a Denison sampler in which the inner barrel is spring loaded so as to 

provide for the automatic adjustment of the distance by which the cutting edge of the barrel leads 

the coring bit. After cleaning the drill hole, the sampler is lowered to the bottom of the drill hole. 

When the sampler reaches the bottom of the drill hole the inner tube meets the resistance first and 

the outer barrel slides past the tube until the spring at the top of the tube contacts the top of the 

outer barrel. The spring in the sampler is compressed with respect to the amount of resistance met 

by the soil sample i.e soft or hard. Sampling is accomplished by rotating the outer barrel at 100 to 

200 rpm while exerting the downward pressure. Upon completion of the sampling drive, the sampler 

is removed from the borehole, and the inner tube which is used to ship and store the sample is 

removed from the sampler. 

Mazier’s Sampler 

The Mazier’s sampler, commonly used in south-east Asia, for soil exploration is very much similar to 

Denison sampler. It is very useful for obtaining samples of stiff to hard residual soil with relict rock 

fragments and weathered material. The Mazier’s triple tube retractor barrel which is a stationary 

plastic liner encasing 73 m diameter core is compatible with standard laboratory and testing 

apparatus. The Mazier’s sample is used in conjunction with double core barrel when coring of rock is 

required. 

Block Samples 

For projects where the determination of the undisturbed properties is very critical, and where the soil 

layers of interest are accessible, undisturbed block samples can be of great value. Of all the 

undisturbed testing methods discussed in this manual, properly-obtained block samples produce 

samples with the least amount of disturbance. Such samples can be obtained from the hillsides, cuts, 

test pits, tunnel walls and other exposed sidewalls. Undisturbed block sampling is limited to cohesive 

soils and rocks. The procedures used for obtaining undisturbed samples vary from cutting large 

blocks of soil using a combination of shovels, hand tools and wire saws, to using small knives and 

spatulas to obtain small blocks. 

In addition, special down-hole block sampling methods have been developed to better obtain 

samples in their in-situ condition. For cohesive soils, the Sherbrooke sampler has been developed 

and is able to obtain samples 250 mm (9.85 in) diameter and 350 mm (13.78 in) height (Lefebvre 

March 2009 2-3


and Poulin 1979). In-situ freezing methods for saturated granular soils and resin impregnation 

methods have been implemented to “lock” the soil in the in-situ condition prior to sampling. When 

implemented, these methods have been shown to produce high quality undisturbed samples. 

However, the methods are rather involved and time consuming and therefore have not seen 

widespread use in practice. 

Once samples are obtained and transported to the laboratory in suitable containers, they are 

trimmed to appropriate size and shape for testing. Block samples should be wrapped with a 

household plastic membrane and heavy duty foil and stored in block form and only trimmed shortly 

before testing. Every sample must be identified with the following information: project number, 

boring or exploration pit number, sample number, sample depth, and orientation. 

2.2.2 Disturbed Sampling 

Disturbed samples are those obtained using equipment that destroy the macrostructure of the soil 

but do not alter its mineralogical composition. Specimens from these samples can be used for 

determining the general lithology of soil deposits, for identification of soil components and general 

classification purposes, for determining grain size, Atterberg limits, and compaction characteristics of 

soils. Disturbed samples can be obtained with a number of different methods as summarized in Table 

2.1. Some of the sampling methods given in Table 2.1 are described below. 

Split-Barrel (Split Spoon) 

The split spoon sampler is a solid steel tube barrel split into two halves longitudinally. The device has 

a check valve and a hard steel shoe. When the head and shoe are unscrewed the barrel opens in the 

centre exposing the sample. Improvement in design provides liner and the sampler retainer. The ball 

valve in the head and the sample retainer valve spring prevent the sample particularly cohesionless 

soil from being washed out and lost. The borehole is cleaned before lowering the sampler into the 

borehole. The sampler is then driven into the borehole base by hammering to extract the sample. 

The sample is then logged on a borelog. 

Continuous Auger 

Continuous auger or continuous flight augers are augers with continuous spiral on the shaft. As the 

hole advances, additional sections of spiral flight are added. In this type of auger, the soils rise to 

the top of the hole on the spiral flight and is sampled as it emerges. Moreover the disadvantage of 

raising and lowering the auger to remove the soil is eliminated. Condinuous augers can be with solid 

or hollow stems also. The limitation of the augers is that these are not effective below water table 

as there are constant caving problems and samples are washed off unless cased. Hollow stem auger 

can cope with the situation to some extent with special adaptors. The limitations are maximum depth 

30m for continuous augers. 

Bulk Samples 

Bulk samples are suitable for soil classification, index testing, R-value, compaction, California Bearing 

Ratio (CBR), and tests used to quantify the properties of compacted geomaterials. The bulk samples 

may be obtained using hand tools without any precautions to minimize sample disturbance. The 

sample may be taken from the base or walls of a test pit or a trench, from drill cuttings, from a hole 

dug with a shovel and other hand tools, by backhoe, or from a stockpile. The sample should be put 

into a container that will retain all of the particle sizes. For large samples, plastic or metal buckets or 

metal barrels are used; for smaller samples, heavy plastic bags that can be sealed to maintain the 

water content of the samples are used. 

2-4 March 2009


Usually, the bulk sample provides representative materials that will serve as borrow for controlled fill 

in construction. Laboratory testing for soil properties will then rely on compacted specimens. If the 

material is relatively homogeneous, then bulk samples may be taken equally well by hand or by 

machine. However, in stratified materials, hand excavation may be required. In the sampling of such 

materials it is necessary to consider the manner in which the material will be excavated for 

construction. If it is likely that the material will be removed layer by layer through the use of 

scrapers, samples of each individual material will be required and hand excavation from base or wall 

of the pit may be a necessity to prevent unwanted mixing of the soils. If, on the other hand, the 

material is to be excavated from a vertical face, then the sampling must be done in a manner that 

will produce a mixture having the same relative amounts of each layer as will be obtained during the 

borrow area excavation. This can usually be accomplished by hand-excavating a shallow trench 

down the walls of the test pit within the depth range of the materials to be mixed. 

2.3 SAMPLING INTERVAL AND APPROPRIATE SAMPLER TYPE 

In general, SPT samples are taken in both granular and cohesive soils, and thin-walled tube samples 

are taken in cohesive soils. The sampling interval will vary between individual projects and between 

regions. A common practice is to obtain split barrel samples at 0.75 m (2.5 ft) intervals in the upper 

3 m (10 ft) and at 1.5 m (5 ft) intervals below 3 m (10 ft). In some instances, a greater sample 

interval, often 3 m (10 ft), is allowed below depths of 30 m (100 ft). In other cases, continuous 

samples may be required for some portion of the boring. 

In cohesive soils, at least one undisturbed soil sample should be obtained from each different 

stratum encountered. If a uniform cohesive soil deposit extends for a considerable depth, additional 

undisturbed samples are commonly obtained at 3 m (10 ft) to 6 m (10 ft) intervals. 

Where borings are widely spaced, it may be appropriate to obtain undisturbed samples in each 

boring; however, for closely spaced borings, or in deposits which are generally uniform in lateral 

extent, undisturbed samples are commonly obtained only in selected borings. In erratic geologic 

formations or thin clay layers it is sometimes necessary to drill a separate boring adjacent to a 

previously completed boring to obtain an undisturbed sample from a specific depth which may have 

been missed in the first boring. 

2.4 SAMPLE RECOVERY 

Occasionally, sampling is attempted and little or no material is recovered. In cases where a split 

barrel or another disturbed-type sample is to be obtained, it is appropriate to make a second attempt 

to recover the soil sample immediately following the first failed attempt. In such instances, the 

sampling device is often modified to include a retainer basket, a hinged trap valve, or other 

measures to help retain the material within the sampler. 

In cases where an undisturbed sample is desired, the field supervisor should direct the driller to drill 

to the bottom of the attempted sampling interval and repeat the sampling attempt. The method of 

sampling should be reviewed, and the sampling equipment should be checked to understand why no 

sample was recovered (such as a plugged ball valve). It may be appropriate to change the sampling 

method and/or the sampling equipment, such as waiting a longer period of time before extracting 

the sampler, extracting the sampler more slowly and with greater care, etc. This process should be 

repeated or a second boring may be advanced to obtain a sample at the same depth. 

2.5 REQUIRED VOLUME OF MATERIAL FOR TESTING PROGRAMME 

A further consideration in fixing sample sizes is the standard test specimen sizes in use. The 

specimen sizes commonly used here and in United Kingdom is shown below. 

March 2009 2-5


Compressibility characteristics 

Oedometer 

Triaxial cell 

Hydraulic consolidation cell 

Triaxial compression tests 

Small specimens 

Large specimens 

Direct shear tests 



76mm dia. x 19mm high 


up to 254mm dia. x 100 – 125mm high 



or 152mm dia. x 305mm high 

60mm x 60mm in plan 

305mm x 305mm in plan 

Small triaxial specimens are normally tested in groups of three, all of which should be obtained from 

the same level in the sample in order that they are as similar as possible. Three 38mm dia. 

Specimens can be obtained from a 102 mm dia. sample. 

Soil testing equipment manufactured in the USA uses the following specimen sizes. 

Compressibility characteristics 

Consolidometer 

(large specimen) 

(small specimen) 

Triaxial compression tests 


Medium specimens 


Direct shear tests 

Cylindrical specimens 

Square specimens 

113mm dia. 

64mm dia. 


71mm dia x 142mm high 


Or 152mm dia. x 305mm high 

63.5mm dia. 

51mm dia. x 52mm 

Three 36mm dia. (1.4in. dia.) specimens can be obtained from either 89mm (3.5 in.) dia. samples or 

102 mm (4 in.) dia. samples. 

As noted above, when discussing the need for samples to contain representative particle sizes, in 

many cases it is the minimum quantity of soil required for a particular test procedure which will 

dictate the volume or mass that must be obtained. BS 5930: 1981 suggested sample sizes should be 

determined on the basis both of soil type and the purpose for which the sample was needed (Table 

2.2). 

2-6 March 2009


Testing 

Table 2.2 Mass of Disturbed Soil Sample Required For Various Tests 

Clay, silt or sand 

(kg) 

Soil type 

Fine and medium 

gravel 

(kg) 

Coarse gravel 

(kg) 

Moisture content, 

Atterberg limits, sieve 

1 5 30 

analysis, chemical tests 

Compaction tests 25-60 25-60 25-60 

Soil stabilization tests 100 130 160 

(Source: BS 5930: 1981) 

2.6 SAMPLE DISTURBANCE 

The most obvious effect of sample disturbance can be seen when attempting to tube sample very 

soft, sensitive clays with a poorly designed sampler. The soil around the edge of the sample 

undergoes a very large decrease in strength, such that when the tube is withdrawn from the soil 

there is no recovery. But, as has been noted above, sample disturbance occurs in all sampling 

processes and, if sampling is carried out well, the effects of disturbance will hopefully be more 

subtle. Whatever its magnitude, sampling disturbance normally affects both undrained strength and 

compressibility. In addition, chemical effects may cause changes in the plasticity and sensitivity of 

the soil sample. 

(I) 

Strength 

Although it has been noted above that tube sampling disturbance has the greatest effect, in terms of 

reductions in mean effective stress, on reconstituted clays its effect on the undrained shear strength 

of such material is, perhaps surprisingly, small. Laboratory experiments by a number of workers have 

shown that the stress paths during undrained shearing converge on the critical state and, because 

the soil is initially reconstituted, the state boundary surface is not disrupted by tube sampling. 

Typically, it has been found that the undrained strength is reduced by less than 10%, even when the 

material is not reconsolidated back to its initial stress state (for example, Siddique (1990)). 

Tube sampling does, however, have a significant effect on real soils, most of which are either 

bonded (‘structured’), and/or more heavily overconsolidated. Shearing of bonded soils during tube 

sampling can have the effect of progressively destructuring them. Clayton et al. (1992) show 

comparisons of the stress paths taken by soil specimens tube sampled in different ways. Figure 2.1 

shows how tube sampling a lightly overconsolidated natural, structured clay with a standard piston 

sampler leads subsequently to much higher pore pressure generation during undrained shear, with 

the consequence that undrained strength is reduced. Clayton et al. (1992) found that provided tube 

sampling strain excursions were limited to ± 2% and that appropriate stress paths were used to 

reconsolidate the material back to its in situ stress state, the undrained strength of the Bothkennar 

clay would be within ± 10% of its undisturbed value. It is to be expected, however, that much 

greater effects will occur when sensitive clays are sampled. 

Heavily overconsolidated clays often display almost vertical stress paths under undrained shear. An 

increase in the mean effective stress level as a result of tube sampling will result in approximately 

proportional increase in intact strength. Unfortunately, however, this is not the only effect at work. 

Hammering of tubes into stiff clays can cause fracturing, and loosening along fissures, and this may 

lead to a marked reduction in measured undrained strength. 

March 2009 2-7

Chapter 2 SAMPLING AND SAMPLING 

DISTURBANCE 

Figure 2.1 Effects of 

Tube Sampling Disturbance of Lightly Overconsolidated Natural (‘Structured’) 

Clay on: (a) Stress Path and Strength during Undrained Triaxial Compression 

(b) 

One-Dimensional Compressibility during Oedometer Testing 

(II) 

Compressibilit 

ty and Stiffness 

The effects of sampling on compressibility (as measured in the oedometer, for example) are difficult 

to assesss because of bedding effects, particularly in heavily overconsolidated clays. The use of local 

axial strain measurement on triaxial specimens during the past decade has produced new and more 

reliable stiffness dataa than can normally be expected from 

routine one-dimensional consolidation 

tests, It is now known that the measured small-strain stiffnesses of clays, most relevant to many 

geotechnical engineering problems, is for a given clay approximately 

linearly proportional to the 

mean effective stress at the time of measurement. This means that changes in effective stress as 

a 

result of disturbance are directly translated into proportional changes in measured soil stiffness. 

Because of the growing appreciation of the influence of bedding and effective stress changes on 

measured stiffness, it 

has becomee common practice in the 

UK to adopt 

laboratory methods which 

will avoid 

these problems. In heavily overconsolidated clays, small-strain stiffness is often normalized 

with respect to the mean effectivee stress at the start of shear (p’o=(σ’1+σ’2+σ’3)/3). Alternatively, 

the stiffness of bonded soils is perhaps more appropriately normalized with respect to undrained 

2-8 

March 2009

Chapter 2 SAMPLING AND SAMPLING 

DISTURBANCE 

shear strength, although it may be 

difficult to determine the true in situ value of this. In situ stiffness 

can then be recovered 

if p’o(in situ) or cu(in situ) be estimated. In lightly overconsolidated natural 

clay Clayton et al. (1992) have shown, however, that even the careful reestablishment of in situ 

effective stress levels before shearing cannot fully recover the undisturbed stiffness behaviour of the 

soil. 

A 60% reduction in E u / p’ o (measured locally, 

and after re-establishment of in situ 

stresses) was 

obtained for the Bothkennar clay following tube sampling strain excursions of ±2% %, for example. 

The results of a literature survey 

by Hopper (1992) are shown in Fig. 2.2. Here the very severe 

effects of tube sampling (including 

the effects of borehole disturbance, and obtained 

by comparing 

test results from tube samples with 

those on block samples in the same soil type) can be seen. 

Figure 2.2 Influence of Tube 

Sampling Disturbance on 

Undrained Strength and Stiffness 

(From a Survey by Hopper 1992). 

March 2009 

2-9


REFERENCES 

[1] Acker, W. L., III (1974). Basic Procedures for Soil Sampling and Core Drilling, Acker Drill Co. 

Inc., P.O. Box 830, Scranton, PA., 18501. 

[2] American Association of State Highway and Transportation Officials (AASHTO) (1988). 

Manual on Subsurface Investigations, Developed by the Subcommittee on Materials, Washington, 

D.C. 

[3] American Association of State Highway and Transportation Officials (AASHTO). (1995). 

Standard specifications for transportation materials and methods of sampling and testing: part II: 

tests, Sixteenth Edition, Washington, D.C. 

[4] Deere, D. U. (1963). “Technical description of rock cores for engineering purposes.” 

Felmechanik und Ingenieur Geologis, 1 (1), 16-22. 

[5] Ford, P.J., Turina, P.J., and Seely, D.E. (1984). Characterization of hazardous waste sites - a 

methods manual: vol. II, available sampling methods, 2nd Edition, EPA 600/4-84-076 (NTIS PB85- 

521596). Environmental Monitoring Systems Laboratory, Las Vegas, NV. 

[6] Hunt, R. E. (1984). Geotechnical Engineering Investigation Manual, McGraw-Hill Inc., 983 p. 

Hvorslev, M. J. (1948). Subsurface Exploration and Sampling of Soils for Civil Engineering Purposes, 

U.S. Army Corps of Engineers, Waterways Experiment Station, Vicksburg, MS. 

[7] Krebs, R. D., and Walker, E. D. (1971). "Highway materials." Publication 272, Department of 

Civil Engrg., Massachusetts Institute of Technology, McGraw-Hill Company, New York, 107. 

[8] Kulhawy, F.H., Trautmann, C.H., and O’Rourke, T.D. (1991). “The soil-rock boundary: What 

is it and where is it?” Detection of and Construction at the Soil/Rock Interface, GSP No. 28, ASCE, 

Reston/VA, 1-15. 

[9] Kulhawy, F.H. and Phoon, K.K. (1993). “Drilled shaft side resistance in clay soil to rock”, 

Design and Performance of Deep Foundations: Piles & Piers in Soil & Soft Rock, GSP No. 38, ASCE, 




[11] Lowe III, J., and Zaccheo, P.F. (1991). "Subsurface explorations and sampling." Foundation 

Engineering Handbook, H. Y. Fang, ed., Van Nostrand Reinhold, New York, 1-71. 

[12] Lupini, J.F., Skinner, A.E., and Vaughan, P.R. (1981). "The drained residual strength of 

cohesive soils", Geotechnique, Vol. 31 (2), 181-213. 

[13] Lutenegger, A. J., DeGroot, D. J., Mirza, C., and Bozozuk, M. (1995). “Recommended 

guidelines for sealing geotechnical exploratory holes.” FHWA Report 378, Federal Highway 

Administration Washington, D.C. 

[14] NAVFAC, P-418. (1983). "Dewatering and groundwater control." Naval Facilities Engineering 

Command, Department of the Navy; Publication No. TM 5-818-5. 

[15] Powers, J. P. (1992). Construction Dewatering, John Wiley & Sons, Inc., New York. 

2-10 March 2009


[16] U.S. Environmental Protection Agency (EPA). (1991). Description and sampling of 

contaminated soils, (EPA/625/12-9/002; November), Washington, D.C. 




Government Printing Office, Washington, D.C 

March 2009 2-11



2-12 March 2009

CHAPTER 3 IN SITU GEOTECHNICAL TESTING

Chapter 3 IN-SITU GEOTECHNICAL TESTING 





3.1 INTRODUCTION .......................................................................................................... 3-1 

3.1 STANDARD PENETRATION TEST (SPT).......................................................................... 3-1 

3.1.1 Correction Factors for Spt .............................................................................. 3-4 

3.2 CONE PENETRATION TEST (CPT).................................................................................. 3-5 

3.3 FIELD VANE SHEAR TEST (VST)................................................................................... 3-15 

3.4 SUMMARY ON IN-SITU GEOTECHNICAL METHODS ........................................................ 3-20 

3.5 GROUNDWATER INVESTIGATIONS .............................................................................. 3-21 

3.5.1 General ....................................................................................................... 3-21 

3.5.2 Determination of Ground Water Levels and Pressures ..................................... 3-22 

3.5.3 Field Measurement of Permeability ................................................................ 3-22 

REFERENCES ....................................................................................................................... 3-24 

March 2009 3-i


List of Table 


3.1 Comparison between Advantages and Disadvantages of SPT 3-4 

3.2 Comparison between Advantages and Disadvantages in CPT 3-5 

3.3 Diagnostic Features of Soil Type 3-14 

3.4 General Advantages and Disadvantages of VST 3-16 

3.5 Field Methods for Measurement of Permeability 3-23 



3.1 Common In-Situ Tests for Geotechnical Site Characterization of Soils 3-1 

3.2(a) Equipment for the Standard Penetration Test 3-2 

3.3 Ratio of Undrained Shear Strength (Cu) Determined On 100mm Diameter. 3-4 

3.4 Original Dutch Cone and Improved Mechanical Delft Cone (Lousberg Et Al. 1974) 3-6 

3.5 Begemann Mechanical Friction Cone (Left, Fully Closed; Right, Fully Extended) 3-7 

3.6 Electric Friction Cone (Largely After Meigh 1987) 3-8 

3.7 Definition of Cone Area Ratio, Α 3-9 

3.8 Distribution of Excess Pore Pressure over the Cone (Coutts 1986). 3-10 

3.9 Typical Record of a Friction Cone Penetration Test (Te Kamp, 1977, from Meigh, 

1987) 3-12 

3.10 (a) relationship between soil type, cone resistance and local friction (Begemann 

1956); 3-13 

3.11 Ratio of (CPT Qc) (SPT N) as a Function of D50 Particle Size of the Soil (Thorburn, 

1971). 3-14 

3.12 General Test Procedures for the Field Vane in Fine-Grained Soils. 3-16 

3.13 Assumed Geometry of Shear Surface for Conventional Interpretation of the Vane 

Test 3-18 

3.14 Vane Correction Factor (:R) Expressed in Terms of Plasticity Index and Time to 

Failure. 3-20 

3.15 Relevance of In-Situ Tests to Different Soil Types 3-21 

3.16 Field Permeability Test Arrangement for Soil 3-23 

3-ii March 2009



3 IN-SITU GEOTECHNICAL TESTING 

Several in-situ tests define the geostratigraphy and obtain direct measurements of soil properties 

and geotechnical parameters. The common tests include: standard penetration test (SPT), cone 

penetration test (CPT), piezocone test (CPTu), flat dilatometer test (DMT), borehole pressure meter 

test (PMT), and vane shear test (VST). Each test applies different loading schemes to measure the 

corresponding soil response in an attempt to evaluate material characteristics, such as strength 

and/or stiffness. Fig. 3.1 depicts these various devices and simplified procedures in graphical form. 

Details on these tests will be given in the subsequent sections. 

Figure 3.1 Common In-Situ Tests for Geotechnical Site Characterization of Soils 

Boreholes are required for conducting the SPT and normal versions of the PMT and VST. A rotary 

drilling rig and crew are essential for these tests. In the case of the CPT, CPTU, and DMT, no 

boreholes are needed, thus termed direct-push technologies. Specialized versions of the PMT (i.e., 

full-displacement type) and VST can be conducted without boreholes. As such, these may be 

conducted using either standard drill rigs or mobile hydraulic systems (cone trucks) in order to 

directly push the probes to the required test depths. 

A disadvantage of direct-push methods is that hard cemented layers and bedrock will prevent further 

penetration. In such cases, borehole methods prevail as they may advance by coring or non-coring 

techniques. An advantage of direct-push soundings is that no cuttings or spoil are generated. 

3.1 STANDARD PENETRATION TEST (SPT) 

The standard penetration test (SPT) is performed during the advancement of a soil boring to obtain 

an approximate measure of the dynamic soil resistance, as well as a disturbed drive sample (split 

barrel type). The test was introduced by the Raymond Pile Company in 1902 and remains today as 

the most common in-situ test worldwide. 

March 2009 3-1


The SPT 

involves the 

driving of a hollow thick-walled tube into the ground and measuring the 

number of blows to advance the split-barrel sampler a vertical distance of 300 mm ( 1 foot). A drop 

weight system is used 

for the pounding where a 63.5-kg (140-lb) hammer repeatedly falls from 0.76 

m (30 inches) to achieve threee successive increments of 150-mm (6-inches) each. The first 

increment is recordedd as a .seating, while the 

numbers of blows to advance the second and third 

increments are summed to give the N-value ("blow count") or SPT-resistance (reported in blows/0.3 

m or blows per foot). Figs. 3.2 a, b refer. 

The penetration resistance (N) is the number of blows required to drive 

the split spoon for the last 

300mm (1 ft) of penetration. The penetration resistancee during the 

first 150 mm (6 in.) of 

penetration is ignored, because the soil is considered to have been disturbed by the action of boring 

the hole. 

Figure 3.2(a) Equipment for the Standard Penetration Test 

3-2 

March 2009


Figure 3.2(b) Sequence of Driving Split-Barrel Sampler During the Standard Penetration Test 

Correlations between SPT N value and soil (Fig. 3.3 refers) or weak rock properties are wholly 

empirical, and depend upon an international database of information. Because the SPT is not 

completely standardised, these correlations cannot be considered particularly accurate in some 

cases, and it is therefore important that users of the SPT and the data it produces have a good 

appreciation of those factors controlling the test, which are: 

1. Variations in the test apparatus; 

2. The disturbance created by boring the hole; and 

3. The soil into which it is driven. 

March 2009 3-3


10 

8 

Cu/N (kN/m 2 ) 

6 

4 

2 

0 

0 10 20 30 40 50 60 70 

PI % 

Boulder clay 

Laminated clay 

Sunnybrook till 

London clay 

Bracklesham bods 

Oxford clay 

Kimmeridge clay 

Woolwich and Reading clay 

Upper Lias clay 

Keuper marl 

Flints 

Figure 3.3 Ratio of Undrained Shear Strength (Cu) Determined On 100mm Diameter. 

Specimens to SPT N, As a Function of Plasticity (Stroud 1974). 

A comparison between advantages and disadvantages of SPT is summarised in Table 3.1 as follows: 

Table 3.1 Comparison between Advantages and Disadvantages of SPT 

Advantages 

Simple and rugged 

Suitable in many soil types 

Can perform in weak rocks 

Easily available 

Disadvantages 

Disturbed sample (index tests only) 

Crude number for analysis 

Not applicable in soft clays and silts 

High variability and uncertainty 

3.1.1 Correction Factors for Spt 

In recent years, it has become a practice to adjust the N valule of SPT test by a hammer-energy 

ratio or hammer efficiency of 60% and much attention has been given to N values because of its 

use in liquefaction studies. Geotechnical foundation practice and engineering usage based on SPT 

correlations have been developed on the basis of standard-of-practice corresponding to an average 

ER = 60 %. Normally the correction factor used for SPT tests N values is 

Where 

(N 1 ) 60 = N.C N .C E (3.1) 

(N 1 ) 60 = Corrected N Value 

N = 

SPT N count obtained from Testing 

3-4 March 2009


C N . = Depth Correction Factor - (Should not be greater than 1.7) 

= (1/σ’ vo ) 0.5 - (Liao and Whitman 1986) 

= 2.2/ (1.2 + σ’ vo / Pa) - (Kayen et. Al, 1992) 

σ’ vo = Effective overburden pressure (γt - γw).z in tons / sq ft 

Pa = 1 tons/sq. ft (95KN/m2) 

C E = Correction Factor for Energy Ratio of 60%. = ER /60 

ER = Energy Ratio for drill rigs (Table below) 

Country Hammer Releases ER (%) 

USA Safety 2 turns of Rope 55 

Donut 2 turns of Rope 45 

Japan Donut Tombi 78 -85 

Donut 2 turns of Rope 65 – 67 

China Automatic Trip 60 

Donut Manual 55 

UK Automatic Trip 73 

Additional correction has been proposed by (Skempton, 1986, Robertson and Wride, 1998) for 

hammer type (donut and safety), borehole diameter rod lengths and sampler. 

3.2 CONE PENETRATION TEST (CPT) 

The cone penetration test is quickly becoming the most popular type of in-situ test because it is fast, 

economical, and provides continuous profiling of geostratigraphy and soil properties evaluation. 

The CPT can be used in very soft clays to dense sands, yet is not particularly appropriate for gravels 

or rocky terrain. The pros and cons are listed in Table 3.2 below. As the test provides more accurate 

and reliable numbers for analysis, yet no soil sampling, it provides an excellent complement to the 

more conventional soil test boring with SPT measurements. 

Table 3.2 Comparison between Advantages and Disadvantages in CPT 

Advantages 

Fast and continuos profiling 

Economical and productive 

Results not operator-dependent 

Strong theoretical basis in interpretation 

Particularly suitable for soft soils 

Disadvantages 

High capital investment 

Requires skilled operator to run 

Electronic drift, noise and calibration 

No soil samples are obtained 

Unsuitable for gravel or boulder deposits 

except where special rigs are provided and / 

or additional drilling support is available. 

Samples of various cone penetrometers are illustrated in Figs. 3.4, 3.5 and 3.6. 

March 2009 3-5


Figure 3.4 Original Dutch Cone and Improved Mechanical Delft Cone (Lousberg Et 

Al. 1974) 

3-6 

March 2009


Figure 3.5 Begemann Mechanical Friction Cone (Left, Fully Closed; Right, Fully Extended) 

(Meigh 1987) 

March 2009 

3-7


Figure 3.6 Electric Friction Cone (Largely After Meigh 

1987) 

Interpretation and use 

The basic 

measurements made by a cone are: 

1. The 

axial force necessary to drive the 10 cm 2 cone into 

the ground at constant velocity; and 

2. The 

axial force generated by 

adhesion or 

friction acting over the 150 cm 2 areaa of the friction 

jacket. 

For piezocones, the basic measurement is the 

pore pressure developed as penetration proceeds. 

Routine calculations convert these measurements into cone resistance, local side friction and friction 

ratio. 

3-8 

March 2009


Cone resistance, q c (normally in MPa) can be calculated from: 

(3.2) 

where F 10cm 2 c = force required to push the cone 

. 

into the ground, and A c 

plan area of 

the cone, i. .e. 

Local side friction, f s (normally in MPa), can be calculated from: 

(3.3) 

where F s shear force on the friction 

sleeve, and A s = area of the friction sleeve, i.e. 150 cm 2 . 

Friction ratio, R f (in %), can be calculated from: 

(3.4) 

Because of the geometry of the electric cone, 

where pore water pressure acts downwards on the 

back of the cone end 

(Fig. 3.7), the cone resistance will be under- recorded. When 

used in deep 

water, for example, for offshore investigations, 

the force exerted by groundwater will be significant, 

and if pore pressuress are measured (with the piezocone), cone resistance can be corrected for this 

effect. The corrected, ‘total’, cone resistance, q t is: 

q t 

where α 

typically 

t = q c +(1- )u 

= ratio of the area of the shaft above the cone end to the area of the c 

0.15 to 0.3, and u = pore pressure at the top of the 

cone. 

(3.5) 

2 ), 

cone (10 cm 2 3-9 

Figure 3.7 Definition of Cone Area Ratio, Α 

March 2009 

3


Because the pore pressure is not always measured at the top of the cone, but 

is sometimes 

measured either on the face, or on the shoulder, a factor must be applied to the measured pore 

pressure. This factor (β) is based upon pore pressure distributions calculated using the 

strain path 

method. Thus: 

qt = q c +(1- )(u 0 +ß∆u) 

(3.6) 

where β = ratio between the calculated excess pore pressure at the top of the cone and at the point 

of measurement, u 0 = hydrostatic 

pore pressure, and ∆u = excess pore pressure caused by cone 

penetration. Pore pressure distributions measured and calculated around piezocones 

are shown in 

Fig. 3.8. 

Figure 3.8 

Distribution 

of Excess Pore Pressure over the Cone 

(Coutts 1986). 

In soft cohesive soils, at depth, much of the cone resistance may be derived from 

the effect of 

overburden, rather than the strength of the soil. In these circumstances the ‘net cone resistance’ 

may be calculated: 

qn = q c -σ v 

(3.7) 

3-10 

March 2009


where q n = net cone resistance, and σ v = vertical total stress at the level at which q n is measured. 

Net cone resistance can only be calculated once the distribution of bulk unit weight with depth is 

known, or can be estimated. 

Typical results of a friction cone test are given in Fig. 3.9. The original development of side friction 

measurement was made by Begemann using a mechanical cone, who found the useful correlation 

between friction ratio and soil type shown in Fig. 3.10a. He defined soil type by its percentage of 

particles finer than 0.016mm, and found that on a plot of side friction versus cone resistance each 

type of soil plotted as a straight line passing through the origin. This has led to more sophisticated 

charts such as that shown in Fig. 3.10b, and for the piezocone to correlations based upon the 

relationship between excess pore pressure and net cone resistance (q n = q c - σ v ). 

March 2009 3-11


Figure 3.9 Typical Record of a Friction Cone Penetration Test (Te Kamp, 1977, From Meigh, 1987) 

3-12 

March 2009


Point resistance, (MPa) 

Cone resistance (kg/cm 2 ) 

Local friction (kg/cm 2 ) 

Friction ratio, P H (%) 

Figure 3.10 (a) relationship between soil type, cone resistance and local friction (Begemann 1956) ); 

(b) Soil identification chart for a mechanical friction cone (Searle 1979) 

The classification of soils is normally carried out on the basis of the value of cone resistance in 

combination with the friction ratio. Generally, the diagnostic features of the common soil types are as 

given in Table 3.3. 

March 2009 

3-13


Table 

3.3 Diagnostic Features of Soil Type 

Soil type 

Organic 

soil 

Normally consolidated clay 

Sand 

Gravel 

Cone resistance 

Low 

Low 

High 

Very high 

Friction ratio 

Very high 

High 

Low 

Low 

Excess pore 

pressure 

Low 

High 

Zero 

Zero 

Useful relationships between angle 

of shearing 

resistance and cone resistance, qc, can be found in 

Schmertmann (1978), and Durgunoglu and Mitchell (1975). A correlation between qc and SPT N, 

based on 

particle size, is shown in Fig. 3.11. 

Figure 

3.11 Ratio of (CPT Qc) (SPT N) As a Function Of D50 Particle Size Of The Soil (Thorburn, 

1971). 

Well-known methods of predicting 

the settlement of shallow footings (de Beer and Martens 1957; 

Schmertmann 1970; Schmertmann et al. 1978) use cone resistance directly. For example, 

Schmertmann et al. ( 1978) use E = 2.5 q c . Such relationships, although of great practical value, are 

known to 

be of limited accuracy. This is to be expected, because the CPT 

test involves the continual 

failure of 

soil around the cone, and cone resistance is a measure of the 

strength of the soil, rather 

than its compressibilit 

ty. 

It has been shown (Lambrechts and Leonards 1978) that while the compressibility of granular soil is 

very significantly affected by over-consolidation, strength is 

not. This shortcoming is 

shared by the 

SPT. However, settlements of spread footings predicted using the CPT tends to be considerably more 

accurate than those using the SPT, 

because there is no borehole disturbance. 

3-14 

March 2009


In a comparative study based upon case records, Dikran found that the ratio of calculated/observed 

settlements fell in the range 0.21—2.72, for four traditional methods of calculation using the CPT. 

For the SPT the variation was 0.15—10.8. 

When calculating the point resistance of piles in sand based upon cone resistance, it is normal to 

consider the static cone penetrometer as a model of the pile, and simply apply a reduction factor of 

between two and six to give allowable bearing pressure (Van der Veen and Boersma 1957; Sanglerat 

1972). Sand deposits are rarely uniform, and so an averaging procedure is used with the q c values 

immediately above and below the proposed pile tip position (Schmertmann 1978). The side friction 

of piles may be calculated directly from the side friction of the cone, or by correlation with cone 

resistance. 

In cohesive soils, the CPT is routinely used to determine both undrained shear strength and 

compressibility. In a similar way to the bearing capacity of a foundation, cone resistance is a function 

of both overburden pressure (σ v ) and undrained shear strength (c u ): 

q c = N k C u +σ v (3.8) 

so that the undrained shear strength may be calculated from: 

c u = q c -σ v 

N k 

(3.9) 

provided that N k is known, or can be estimated. The theoretical bearing capacity factor for deep 

foundation failure cannot be applied in this equation because the cone shears the soil more rapidly 

than other tests, and the soil is failed very much more quickly than in a field situation such as an 

embankment failure. 

At shallow depths, or in heavily over-consolidated soils, the vertical total stress in the soil is small, so 

that: 

c u q c 

N k 

(3.10) 

Typically, in these conditions, the undrained shear strength is about 1/15th to 1/20th of the cone 

resistance. 

N k is not a constant, but depends upon cone type, soil type, overconsolidation ratio, degree of 

cementing, and the method by which undrained shear strength has been measured (because 

undrained shear strength is sample-size and test-method dependent). The N k value in an overconsolidated 

clay will be higher than in the same clay when normally consolidated 

Typically, N k varies from 10 to 20. Lunne and Kleven have shown that this variation is significantly 

reduced, giving N k much closer on average to 15, if a correction (N k * = N k /µ) is made to allow for 

rate effects, in a similar way to that proposed by Bjerrum for the vane test (see below), but this is 

rarely done in practice. 

3.3 FIELD VANE SHEAR TEST (VST) 

The vane shear test (VST), or field vane (FV), is used to evaluate the in-place undrained shear 

strength (s uv ) of soft to stiff clays & silts at regular depth intervals of 1 meter (3.28 feet). The test 

consists of inserting a four-bladed vane into the clay and rotating the device about a vertical axis. 

Limit equilibrium analysis is used to relate the measured peak torque to the calculated value of s u . 

Both the peak and remoulded strengths can be measured; their ratio is termed the sensitivity, S t . A 

March 2009 3-15


selection of vanes is available in terms of size, shape, and configuration, depending upon the 

consistency and strength characteristics of the soil. The standard vane has a rectangular geometry 

with a blade diameter D = 65 mm, height H = 130 mm (H/D =2), and blade thickness e = 2 mm. 

Fig. 3.12 illustrates the general VST procedures 

Figure 3.12 General Test Procedures for the Field Vane in Fine-Grained Soils. (Note: Interpretation of 

Undrained Strength Shown Is Specifically For Standard Rectangular Vane with H/D = 2) 

The general advantages and disadvantages of VST is summarised in Table 3.4 as follows. 

Table 3.4 General Advantages and Disadvantages of VST 

Advantages 

Assessment of undrained strength, s uv 

Simple test and equipment 

Measure in-situ clay sensitivity (S t ) 

Long history of use in practice 

Disadvantage 

Limited application to soft to stiff clays 

Slow and time-consuming 

Raw s uv needs (empirical) correction 

Can be affected by sand lenses and seams 

By implication, BS 1377 considers that the field vane will not be suitable for testing soils with 

undrained strengths greater than about 75 kPa. The vane must be designed to achieve an area ratio 

of 12% or less. The test is not suitable for fibrous peats, sands or gravels, or in clays containing 

laminations of silt or sand, or stones. 

3-16 March 2009


Interpretation 

The vane test is routinely used only to obtain ‘undisturbed’ peak undrained shear strength, and 

remoulded undrained shear strength. The undrained strength is derived on the basis of the following 

assumptions: 

1. Penetration of the vane causes negligible disturbance, both in terms of changes in effective 

stress, and shear distortion; 

2. No drainage occurs before or during shear; 

3. The soil is isotropic and homogeneous; 

4. The soil fails on a cylindrical shear surface; 

5. The diameter of the shear surface is equal to the width of the vane blades; 

6. At peak and remoulded strength there is a uniform shear stress distribution across the shear 

surface; and 

7. There is no progressive failure, so that at maximum torque the shear stress at all points on 

the shear surface is equal to the undrained shear strength, c u . 

On this basis (Fig. 3.13), the maximum torque will be: 

T = D2 Hc u 

2 

D/2 

+ 2 2δr-rc 

0 

u 

(3.11) 

= D2 Hc u 

2 

= D2 H 

2 

+ 4πr3 D/2 

3 c u0 

1+ D 

3H c u 

For a vane blade where H = 2D: 

T = 3.667D 3 c u (3.12) 

March 2009 3-17


Figure 3.13 Assumed Geometry of Shear Surface for Conventional Interpretation of the Vane Test 

If it is assumed that the shear stress mobilized by the soil is linearly proportional to displacement, up 

to failure, then another simple assumption (Skempton 1948), that the shear stress on the top and 

bottom of the cylindrical shear surface has a triangular distribution, is sometimes adopted. For the 

rectangular vane this leads to the equation: 

T = D2 H 

1+ D 2 4H c u (3.13) 

For a vane blade where H = 2D: 

T = 3.53D 3 c u (3.14) 

giving only 4% difference in shear strength from that obtained using the uniform assumption. 

Undrained Strength and Sensitivity 

The conventional interpretation for obtaining the undrained shear strength from the recorded 

maximum torque (T) assumes a uniform distribution of shear stresses both top and bottom along the 

blades and a vane with height-to-width ratio H/D = 2 (Chandler, 1988), as given in Eq. 3-11 above, 

regardless of units so long as torque T and width D are in consistent units (e.g., kN-m and meters, 

respectively, to provide vane strength c uv in kN/m 2 ). The test is normally reserved for soft to stiff 

materials with c uv < 200 kPa. (2 tsf). After the peak c uv is obtained, the vane is rotated quickly 

through 10 complete revolutions and the remoulded (or "residual") value is recorded. The in-situ 

sensitivity of the soil is defined by: 

3-18 March 2009


S t = c u(peak) /c u(remolded) (3.15) 

For the commercial vanes in common use, the following expressions for vanes with blade heights 

that are twice their widths (H/D = 2) are obtained: 

Rectangular (i T = 0° and i B = 0°): s uv = 0.273 T max /D 3 (3.16a) 

Nilcon (i T = 0° and i B = 45°): s uv = 0.265 T max /D 3 (3.16b) 

Geonor (i T = 45° and i B = 45°): s uv = 0.257 T max /D 3 (3.16c) 

Vane Correction Factor 

It is very important that the measured vane strength be corrected prior to use in stability analyses 

involving embankments on soft ground, bearing capacity, and excavations in soft clays. The 

mobilized shear strength is given by: 

τ mobilized = μ R s uv (3.17) 

where μ R = empirical correction factor that has been related to plasticity index (PI) and/or liquid 

limit (LL) based on back-calculation from failure case history records of full-scale projects. An 

extensive review of the factors and relationships affecting vane measurements in clays and silts with 

PI > 5% recommends the following expression (Chandler, 1988): 

μ R = 1.05 - b (PI) 0.5 (3.18) 

where the parameter b is a rate factor that depends upon the time-to-failure (t f in minutes) and 

given by: 

b = 0.015 + 0.0075 log t f (3.19) 

The combined relationships are shown in Fig. 3.14. For guidance, embankments on soft ground are 

normally associated with t f on the order of 10 4 minutes because of the time involved in construction 

using large equipment. 

March 2009 3-19


Figure 3.14 Vane Correction Factor (:R) Expressed in Terms of Plasticity Index and Time to Failure. 

(Adapted from Chandler, 1988). Note: For Stability Analyses Involving Normal Rates of Embankment 

Construction, the Correction Factor is Taken at the Curve Corresponding to T f = 10,000 Minutes. 

It has been shown that the mobilized undrained shear strength back-calculated from failure case 

histories involving embankments, foundations, and excavations in soft clays are essentially 

independent of plasticity index (Terzaghi, et al. 1996). 

For further information, a detailed review of the device, the procedures, and methods of 

interpretation for the VST are given by Chandler (1988). 

3.4 SUMMARY ON IN-SITU GEOTECHNICAL METHODS 

In-situ physical testing provide direct information concerning the subsurface conditions, geostratigraphy, 

and engineering properties prior to design, bids, and construction on the ground. 

In soils, in-situ geotechnical tests include penetration-type (Standard Penetration Test (SPT), Cone 

Penetration Test (CPT), Cone Penetrometer Test / Piezocone Test (CPTu), Flat Dilatometer Test 

(DMT), Cone Pressuremeter (CPMT), Vane Shear Test (VST)) and probing-type (Pressuremeter Test 

(PMT), Self-boring Pressurementer(SBP)) methods to directly obtain the response of the 

geomaterials under various loading situations and drainage conditions. 

The general applicability of the test method depends in part on the geo-material types encountered 

during the site investigation, as shown by Figure 3.15. The relevance of each test also depends on 

the project type and its requirements. 

Commonly used penetration type tests are Standard Penetration Test (SPT), Cone Penetration Test 

(CPT) and Vane Shear Test (VST). Other tests are carried out for special purposes and requirements. 

3-20 March 2009


SPT 

In-situ Test Method 

CPT 

DMT 

PMT 

VST 

Geophysics 

Grain size (mm) 

Figure 3.15 Relevance of In-Situ Tests to Different Soil Types 

The evaluation of strength, deformation, flow, and time-rate behaviour of soil materials can be 

derived from selected tests or combinations of these test methods. Together, information from these 

tests allow for the rational and economical selection for deciding foundation types for bridges and 

buildings, safe embankment construction over soft ground, cut angles for adequate slope stability, 

and lateral support for underground excavations. 

3.5 GROUNDWATER INVESTIGATIONS 

3.5.1 General 

Groundwater conditions and the potential for groundwater seepage are fundamental factors in 

virtually all geotechnical analyses and design studies. Accordingly, the evaluation of groundwater 

conditions is a basic element of almost all geotechnical investigation programs. Groundwater 

investigations are of two types as follows: 

o 

o 

Determination of groundwater levels and pressures and 

Measurement of the permeability of the subsurface materials. 

Determination of groundwater levels and pressures includes measurements of the elevation of the 

groundwater surface or water table and its variation with the season of the year; the location of 

perched water tables; the location of aquifers (geological units which yield economically significant 

amounts of water to a well); and the presence of artesian pressures. Water levels and pressures may 

be measured in existing wells, in boreholes and in specially-installed observation wells. Piezometers 

are used where the measurement of the ground water pressures are specifically required (i.e. to 

determine excess hydrostatic pressures, or the progress of primary consolidation). 

Determination of the permeability of soil or rock strata is needed in connection with surface water 

and groundwater studies involving seepage through earth dams, yield of wells, infiltration, 

excavations and basements, construction dewatering, contaminant migration from hazardous waste 

March 2009 3-21


spills, landfill assessment, and other problems involving flow. Permeability is determined by means of 

various types of seepage, pressure, pumping, and flow tests. 

3.5.2 Determination of Ground Water Levels and Pressures 

Observations of the groundwater level and pressure are an important part of all geotechnical 

explorations, and the identification of groundwater conditions should receive the same level of care 

given to soil descriptions and samples. Measurements of water entry during drilling and 

measurements of the groundwater level at least once following drilling should be considered a 

minimum effort to obtain water level data, unless alternate methods, such as installation of 

observation wells, are defined by the geotechnical engineer. 

3.5.3 Field Measurement of Permeability 

The permeability (k) is a measure of how easily water and other fluids are transmitted through the 

geo-material and thus represents a flow property. In addition to groundwater related issues, it is of 

particular concern in geo-environmental problems. The parameter k is closely related to the 

coefficient of consolidation (c v ) since time rate of settlement is controlled by the permeability. In 

geotechnical engineering, we designate small k = coefficient of permeability or hydraulic conductivity 

(units of cm/sec), which follows Darcy's law: 

q = kiA (3.20) 

where q = flow (cm 3 /sec), i = hydraulic gradient, and A = cross-sectional area of flow. 

Laboratory permeability tests may be conducted on undisturbed samples of natural soils or rocks, or 

on reconstituted specimens of soil that will be used as controlled fill in embankments and earthen 

dams. Field permeability tests may be conducted on natural soils (and rocks) by a number of 

methods, including simple falling head, packer (pressurized tests), pumping (drawdown), slug tests 

(dynamic impulse), and dissipation tests. A brief listing of the field permeability methods is given in 

Table 3.5. Field permeability arrangement for soil and rock are presented in Figure 3.16 and Figure 

3.17. 

3-22 March 2009


Table 3.5 Field Methods for Measurement of Permeability 

Test Method 

Applicable Soils 

Reference 

Various Field Methods 

Soil and Rock Aquifers 

ASTM D4043 

Pumping tests 

Double-ring 

infiltrometer 

Infiltrometer with 

sealed 

ring 

Various field methods 

Slug tests 

Hydraulic fracturing 

Constant head injection 

Pressure pulse 

technique 

Piezocone dissipation 

Dilatometer dissipation 

Falling 

head tests 

Drawdown in soilss 

Surface fill soils 

Surface soils 

Soils in vadose zone 

Soils at depth 

Rock in-situ 

Low-permeabilitrocks 

Low-permeabilitrocks 

Low to medium k soils 

Low to medium k soils 

Cased 

borehole in soils 

ASTM D4050 

ASTM D3385 

ASTM D5093 

ASTM D5126 

ASTM D4044 

ASTM D4645 

ASTM D4630 

ASTM D4630 

Houlsby & The 

(1988) 

Robertson et al. 

(1988) 

Lambe & Whitman 

(1979) 

BS-5930 –(1988) 

Figure 3.16 Field Permeability Test Arrangement for Soil 

March 2009 

3-23


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seismic piezocone tests”. Geotechnical Site Characterization, Vol. 2, Balkema, Rotterdam, 1113-1118. 

[43] Mayne, P.W. and Martin, G.K. (1998). “Commentary on Marchetti flat dilatometer 

correlations in soils.” ASTM Geotechnical Testing Journal, Vol. 21 (3), 222-239. 

[44] Mayne, P.W., Schneider, J.A., and Martin, G.K. (1999). "Small- and large-strain soil 

properties from seismic flat dilatometer tests", Pre-Failure Deformation Characteristics of 

Geomaterials, Vol. 1 (Torino), Balkema, Rotterdam, 419-426. 

[45] Mayne, P.W. (2001). “Stress-strain-strength-flow parameters from enhanced in-situ tests”, 

Proceedings, International Conference on In-Situ Measurement of Soil Properties & Case Histories 

(In-Situ 2001), Bali, Indonesia, 47-69. 

[46] Parez, L. and Faureil, R. (1988). “Le piézocône. Améliorations apportées à la reconnaissance 

de sols”. Revue Française de Géotech, Vol. 44, 13-27. 

3-26 March 2009


[47] Rehm, B.W., Stolzenburg, T. R., and Nichols, D. G. (1985). “Field measurement methods for 

hydrogeologic investigations: a critical review of the literature.” EPRI Report No. EA-4301, Electric 

Power Research Institute, Palo Alto, CA. 

[48] Robertson, P.K. and Campanella, R.G. (1983). “Interpretation of cone penetration tests: Part 

I - sands; Part II - clays”. Canadian Geotechnical Journal, Vol. 20 (4), 719-745. 

[49] Robertson, P.K., Campanella, R.G., and Wightman, A. (1983). “SPT-CPT correlations”, 

Journal of the Geotechnical Engineering Division (ASCE), Vol. 109 (11), 1449-1459. 

[50] Robertson, P.K. (1986). “In-situ testing and its application to foundation engineering”, 

Canadian Geotechnical Journal, Vol. 23 (4), 573-584. 

[51] Robertson, P.K., Campanella, R.G., Gillespie, D., and Grieg, J. (1986). “Use of piezometer 

cone data”. Use of In-Situ Tests in Geotechnical Engineering, GSP No. 6, ASCE, New York, 1263- 

1280. 

[52] Robertson, P.K., Campanella, R.G., Gillespie, D., and Rice, A. (1986). “Seismic CPT to 

measure in-situ shear wave velocity”. Journal of Geotechnical Engineering 112 (8), 71-803. 

[53] Robertson, P.K., Campanella, R.G., Gillespie, D. and By, T. (1988). “Excess pore pressures 

and the flat dilatometer test”, Penetration Testing 1988, Vol. 1, Balkema, Rotterdam, 567-576. 

[54] Robertson, P.K. (1990). “Soil classification using the cone penetration test”. Canadian 

Geotechnical Journal, Vol. 27 (1), 151-158. 

[55] Santamarina, J.C., Klein, K. and Fam, M.A. (2001). Soils and Waves, Particulate Materials 

Behavior, Characterization, & Process Monitoring, John Wiley & Sons, Ltd., New York, 488 p. 

[56] Schmertmann, J.H. (1986). “Suggested method for performing the flat dilatometer test”, 

ASTM Geotechnical Testing Journal, Vol. 9 (2), 93-101. 

[57] Skempton, A.W. (1986). “SPT procedures and the effects in sands of overburden pressure, 

relative density, particle size, aging, and overconsolidation”. Geotechnique, Vol. 36, No. 3, 425-447. 

[58] Stokoe, K. H., and Woods, R. D. (1972). "In-situ shear wave velocity by cross-hole method." 

Journal of the. Soil Mechanics &.Foundations Division, ASCE, 98 (5), 443-460. 

[59] Stokoe, K. H., and Hoar, R. J. (1978). "Variables affecting in-situ seismic measurement." 

Proceedings, Earthquake Engineering and Soil Dynamics, ASCE, Pasadena, Ca, 919-938. 

[60] Tanaka, H. and Tanaka, M. (1998). "Characterization of sandy soils using CPT and DMT", 

Soils and Foundations, Vol. 38 (3), 55-67 

[61] Tatsuoka, F. and Shibuya, S. (1992). “Deformation characteristics of soils & rocks from field 

& lab tests.” Report of the Institute of Industrial Science 37 (1), Serial No. 235, University of Tokyo, 

136 p. 

[62] Tavenas, F., LeBlond, P., Jean, P., and Leroueil, S. (1983). “The permeability of natural soft 

clays: Parts I and II”, Canadian Geotechnical Journal, Vol. 20 (4), 629-660. 

March 2009 3-27


[63] Teh, C.I. and Houlsby, G.T. (1991). “An analytical study of the cone penetration test in clay”. 

Geotechnique, Vol. 41 (1), 17-34. 



[65] U.S. Army Corps of Engineers. (1951). "Time lag and soil permeability in groundwater 

observations." Waterways Experiment Station, Bulletin No. 36, Vicksburg, MS. 



[67] Windle, D., and Wroth, C. P. (1977). "In-situ measurement of the properties of stiff clays." 

Proceedings, 9th International Conference on Soil Mechanics and Foundation Engineering, Vol. 1, 

Tokyo, Japan, 347-352. 

3-28 March 2009

CHAPTER 4 LAB TESTING FOR SOILS

z 

Chapter 4 LABORATORY TESTING FOR SOILS 





4.1 GENERAL .................................................................................................................... 4-1 

4.2 WEIGHT – VOLUME CONCEPTS .................................................................................... 4-1 

4.3 LOAD-DEFORMATION PROCESS IN SOILS ..................................................................... 4-2 

4.4 PRINCIPLES OF EFFECTIVE STRESS .............................................................................. 4-3 

4.5 OVERBURDEN STRESS ................................................................................................. 4-3 

4.6 TESTS FOR GEOTECHNICAL PARAMETERS .................................................................... 4-4 

4.6.1 Classification Tests ........................................................................................ 4-5 

4.6.2 Chemical and Electro-chemical Tests .............................................................. 4-7 

4.6.3 Compaction Related Tests .............................................................................. 4-8 

4.6.4 Compressibility, Permeability and Durability Tests ............................................ 4-9 

4.6.5 Consolidation and Permeability Tests in Hydraulic Cells and with Pore Pressure 

Measurement ................................................................................................ 4-9 

4.6.6 Shear Strength Tests (Total and Effective Stresses) ........................................ 4-10 

REFERENCES ....................................................................................................................... 4-16 

March 2009 4-i


List of Table 


4.1 Terms in Weight – Volume Relations (After Cheney And Chassie, 1993) 4-1 

4.2 Unit Weight – Volume Relationships 4-2 

4.3 Available Chemical Tests 4-7 



4.1 Typical Particle Size Distribution 4-5 

4.2 Casagrande Plot Showing Classification of Soil into Groups 4-7 

4.3 Typical Compaction Curves 4-8 

4.4 Consolidation Test Apparatus 4-10 

4.5 Bishop Direct Shear Box 4-12 

4.6 Triaxial Cell 4-13 

4-ii March 2009


4 LABORATORY TESTING FOR SOILS 

4.1 GENERAL 

Laboratory testing of soils is a fundamental element of geotechnical engineering. The complexity of 

testing required for a particular project may range from a simple moisture content determination to 

specialized strength and stiffness testing. Since testing can be expensive and time consuming, the 

geotechnical engineer should recognize the projects issues ahead of time so as to optimize the 

testing program, particularly strength and consolidation testing. 

Before describing the various soil test methods, soil behavioural under load will be examined and 

common soil mechanics terms introduced. The following discussion includes only basic concepts of 

soil behaviour. However, the engineer must grasp these concepts in order to select the appropriate 

tests to model the in-situ conditions. The terms and symbols shown will be used in all the remaining 

modules of the course. Basic soil mechanics textbooks should be consulted for further explanation of 

these and other terms. 

4.2 WEIGHT – VOLUME CONCEPTS 

A sample of soil is usually composed of soil grains, water and air. The soil grains are irregularly 

shaped solids which are in contact with other adjacent soil grains. The weight and volume of a soil 

sample depends on the specific gravity of the soil grains (solids), the size of the space between soil 

grains (voids and pores) and the amount of void space filled with water. Common terms associated 

with weight-volume relationships are shown in Table 4.1. 

1 

Table 4.1 Terms in Weight – Volume Relations (After Cheney And Chassie, 1993) 

Property Symbol Units 1 How obtained 

(AASHTO/ASTM/BSS) 

Moisture Content w D 

By measurement 

(T 265/D 4959/BS1377-Part 2) 

Specific Gravity G c D 

By measurement 

(T 100/D 854 BS1377-Part 2) 

Unit weight FL -3 By measurement or from 

weight-volume relations 

Direct Applications 

Classification and in 

weight-volume relations 

Volume computations 

Classification and for 

pressure computations 

Porosity n D From weight-volume relations 

Defines relative volume 

of solids to total volume 

of soil 

Void Ratio e D From weight-volume relations 

Defines relative volume 

of solids to total volume 

of soil 

F = Force or weight; L = Length; D = Dimensionless. Although by definition, moisture content is 

a dimensionless fraction (ratio of weight of water of solids), it is commonly reported in percent 

by multiplying the fraction by 100. 

Of particular note is the void ratio (e) which is a general indicator of the relative strength and 

compressibility of a soil sample, i.e., low void ratios generally indicates strong soils of low 

compressibility, while high void ratios are often indicative of weak and highly compressible soils. 

Selected weight-volume (unit weight) relations are presented in Table 4.2. 

March 2009 4-1


Table 4.2 Unit Weight – Volume Relationships 

Case Relationship Applicable Geomaterials 

Soil Identities 

1. G δ w = S e 

All types of soils and rocks 

2. Total Unit Weight: 

= 1+w 

1+e G s w 

Limiting Unit Weight 

Dry Unit Weight 

Moist Unit Weight 

(Total Unit Weight) 

Saturated Unit Weight 

Solid phase only: w=e=0: 

rock = G s w 

For w=0 (all air in void space): 

d = G s w /(1+e) 

Variable amounts of air and water: 

t = G s w (1+w)/(1+e) 

with e = G δ w/S 

Set S = 1 (all voids with water): 

sat = w (G s +e)/(1+e) 

Maximum expected value for 

solid silica is 27 kN/m 3 

Use for clean sands and dry 

soils above groundwater table 

Partially-saturated soils above 

water table; depends on degree 

of saturation (S, as decimal) 

All soils below water table; 

Saturated clays and silts above 

water table with full capillarity 

Hierarchy d ≤ t ≤ sat < rock Check on relative values 

Note: w = 9.8 kN/m 3 (62.4 pcf) for fresh water 

4.3 LOAD-DEFORMATION PROCESS IN SOILS 

When a load is applied to a soil sample, the deformation which occurs will depend on the grain-tograin 

contact (inter-granular) forces and the amount of water in the voids. If no porewater exists, 

the sample deformation will be due to sliding between soil grains and deformation of the individual 

soil grains. The rearrangement of soil grains due to sliding accounts for most of the deformation. 

Adequate deformation is required to increase the grain contact areas to take the applied load. As the 

amount of pore water in the void increases, the pressure it exerts on soil grains will increase and 

reduce the inter-granular contact forces. 

In fact, tiny clay particles may be forced completely apart by water in the pore space. Deformation of 

a saturated soil is more complicated than that of dry soil as water molecules, which fill the voids, 

must be squeezed out of the sample before readjustment of soil grains can occur. The more 

permeable a soil is, the faster the deformation under load will occur. However, when the load on a 

saturated soil is quickly increased, the increase is carried entirely by the pore water until drainage 

begins. Then more and more load is gradually transferred to the soil grains until the excess pore 

pressure has dissipated and the soil grains readjust to a denser configuration. This process is called 

consolidation and results in a higher unit weight and a decreased void ratio. 

4-2 March 2009


4.4 PRINCIPLES OF EFFECTIVE STRESS 

The consolidation process demonstrates the very important principle of effective stress, which will be 

used in all the remaining modules of this course. 

Under an applied load, the total stress in a saturated soil sample is composed of the inter-granular 

stress and porewater pressure (neutral stress). As the porewater has zero shear strength and is 

considered incompressible, only the inter-granular stress is effective in resisting shear or limiting 

compression of the soil sample. Therefore, the inter-granular contact stress is called the effective 

stress. Simply stated, this fundamental principle states that the effective stress (σ’) on any plane 

within a soil mass is the net difference between the total stress (σ t ) and porewater pressure (u). 

When pore water drains from soil during consolidation, the area of contact between soil grains 

increases, which increases the level of effective stress and therefore the soil’s shear strength. In 

practice, staged construction of embankments is used to permit increase of effective stress in the 

foundation soil before subsequent fill load is added. In such operations the effective stress increase 

is frequently monitored with piezometers to ensure the next stage of embankment can be safely 

placed. 

Soil deposits below the water table will be considered saturated and the ambient pore pressure at 

any depth may be computed by multiplying the unit weight of water (γ w ) by the height of water 

above that depth. For partially saturated soil, the effective stress will be influenced by the soil 

structure and degree of saturation (Bishop, et. al., 1960). In many cases involving silts and clays, the 

continuous void spaces that exist in the soil behave as capillary tubes of variable cross-section. Due 

to capillarity, water may rise above the static groundwater table (phreatic surface) as a negative 

porewater pressure and the soils may be nearly or fully saturated. 

4.5 OVERBURDEN STRESS 

The purpose of laboratory testing is to simulate in-situ soil loading under controlled boundary 

conditions. Soils existing at a depth below the ground surface are affected by the weight of the soil 

above that depth. The influence of this weight, known generally as the overburden stress, causes a 

state of stress to exist which is unique at that depth for that soil. When a soil sample is removed 

from the ground, that state of stress is relieved as all confinement of the sample has been removed. 

In testing, it is important to re-establish the in-situ stress conditions and to study changes in soil 

properties when additional stresses representing the expected design loading are applied. In this 

regard, the effective stress (grain-to-grain contact) is the controlling factor in shear, state of stress, 

consolidation, stiffness, and flow. Therefore, the designer should try to re-establish the effective 

stress condition during most testing. 

The test confining stresses are estimated from the total, hydrostatic, and effective overburden 

stresses. The engineer’s first task is determining these stress and pressure variations with depth. 

This involves determining the total unit weights (density) for each soil layer in the subsurface profile, 

and determining the depth of the water table. Unit weight may be accurately determined from 

density tests on undisturbed samples or estimated from in-situ test measurements. The water table 

is routinely recorded on the boring logs, or can be measured in open standpipes, piezometers, and 

dissipation tests during CPTs and DMTs. 

The total vertical (overburden) stress (σ vo ) at any depth (z) may be found as the accumulation of 

total unit weights ( t ) of the soil strata above that depth: 

σ vo = t dz= ∑ t ∆z (4.1) 

March 2009 4-3


For soils above the phreatic surface, the applicable value of total unit weight may be dry, moist, or 

saturated depending upon the soil type and degree of capillarity (see Table 4.2). For soil elements 

situated below the groundwater table, the saturated unit weight is normally adopted. 

The hydrostatic pressure depends upon the degree of saturation and level of the phreatic surface 

and is determined as follow: 

Soil elements above water table: u o = 0 (Completely dry) (4.2a) 

= w (z-z w ) (Full capillarity) (4.2b) 

Soils elements below water table: u o = w (z-z w ) 

(4.2c) 

where z = depth of soil element, z w = depth to groundwater table. Another case involves partial 

saturation with intermediate values between (4.2a and 4.2b) which literally vary daily with the 

weather and can be obtained via tensiometer measurements in the field. Usual practical calculations 

adopt (4.2a) for many soils, yet the negative capillary values from (4.2b) often apply to saturated 

clay and silt deposits. 

The effective vertical stress is obtained as the difference between (4.1) and (4.2): 

σ vo ’ = σ vo - u o (4.3) 

A plot of effective overburden profile with depth is called a ’ v diagram and is extensively used in all 

aspects of foundation testing and analysis (see Holtz & Kovacs, 1981; Lambe & Whitman, 1979). 

4.6 TESTS FOR GEOTECHNICAL PARAMETERS 

A wide range of tests has been used to determine the geotechnical parameters required in 

calculations for example, of bearing capacity, slope stability, earth pressure and settlement. 

Geotechnical calculations remain almost entirely semi-empirical in nature; it has been said that when 

calculating the stability of a slope one uses the ‘wrong’ slip circle with the ‘wrong’ shear strength to 

arrive at a satisfactory answer. For this reason testing requirements differ considerably from region 

to region. 

The new British Standard (BS 1377:1990.) is divided into nine separate parts: 

Part 1 

Part 2 

Part 3 

Part 4 

Part 5 

Part 6 

Part 7 

Part 8 

Part 9 

General requirements and sample preparation 

Classification tests 

Chemical and electro-chemical tests 

Compaction-related tests 

Compressibility, permeability and durability tests 

Consolidation and permeability tests in hydraulic cells and with pore pressure 

measurement 

Shear strength tests (total stress) 

Shear strength tests (effective stress) 

In situ tests. 

4-4 March 2009

Chapter 4 LABORATORY TESTING 

FOR SOILS 

4.6.1 

Classification Testss 

Soil classification, although introducing a further stage of data acquisition 

into site investigation, has 

an important role to play in reducing the costs and increasing the cost-effectiveness of laboratory 

testing. Together with 

detailed sample description, classification tests allow the soils on a site to be 

divided into a limited number of arbitrary groups, each of which is estimated to contain materials of 

similar geotechnical properties. Subsequent more expensive and time-consuming testss carried out to 

determine geotechnical parameters for design purposes may then be made on limited numbers of 

samples which are selected to be representative 

e of the soil group in question. 

Particle Size Distribution Tests 

BS 1377:1990 gives four methods for determining the particle size distribution of 

soils (part 2, 

clauses 9.2—9.5). The coarse fraction of the soil (>0.06mmm approximately) is tested 

by passing it 

through a series of sieves with diminishing apertures. The particle size distribution is 

obtained from 

records of the weight of soil particles retained 

on each sieve and is usually shown as a graph of 

‘percentage passing by weight’ as a function of particle size (Fig. 4.1). 

Figure 4.1 Typical Particle Size Distribution 

Two methods of sieving are defined in BS 13777 (part 2, clauses 9.2, 9.3) . Dry sieving is only suitable 

for sandss and gravels 

which do not contain any clay: the British Standard discourages its use, and 

since the 

exact composition of a soil will not be 

known before testing, it is not often requested. Wet 

sieving requires a complex procedure to separate the fine clayey particles 

from the coarse fraction of 

the soil which is suitable for sieving, as summarized below. 

1. Select representative test specimen by quartering and 

riffling. 

2. Oven dry specimen at 105— —110°C, and weigh. 

3. Place on 20mmm sieve. 

4. Wirebrush each 

particle retained on the 20mm sieve to remove fines. 

5. Sieve particles coarser than 20 mm. Record weights retained on each sieve. 

6. Riffle particles finer than 20mm to reducee specimen mass to 2kg (approx.) weigh. 

March 2009 

4-5


7. Spread soil in a tray and cover with water and sodium hexametaphosphate (2 g/l). 

8. Stir frequently for 1 h, to break down and separate clay particles. 

9. Place soil in small batches on a 2mm sieve resting on a 63 m sieve and wash gently to remove 

fines. 

10. When clean, place the material retained in an oven and dry at 105—110°C. 

11. Sieve through standard mesh sizes between 20mm and 6.3 mm using the dry sieving 

procedure. Note weights retained on each sieve. 

12. f more than 150 g passes the 6.3mm mesh, split the sample by riffling to give 100—150g. 

13. Sieve through standard mesh sizes between 5mm and 63 tm sieve. 

The particle size distribution of the fine soil fraction, between about 0.1 mm and 1 µm may be 

determined by one of two British Standard sedimentation tests (BS 1377:part 2, clauses 9.4, 9.5). 

Soil is sedimented through water, and Stokes’ law, which relates the terminal velocity of a spherical 

particle falling through a liquid of known viscosity to its diameter and specific gravity, is used to 

deduce the particle size distribution. 

Sedimentation tests make a number of important assumptions. Since Stokes’ law is used, the 

following assumptions are implied (Allen 1975). 

1. The drag force on each particle is due entirely to viscous forces within the fluid. The particles 

must be spherical, smooth and rigid, and there must be no slippage between them and the 

fluid. 

2. Each particle must move as if it were a single particle in a fluid of infinite extent. 

3. The terminal velocity must be reached very shortly after the test starts. 

4. The settling velocity must be slow enough so that inertia effects are negligible. 

5. The fluid must be homogeneous compared with the size of the particle. 

Plasticity tests 

The plasticity of soils is determined by using relatively simple remoulded strength tests. The plastic 

limit is the moisture content of the soil under test when remoulded and rolled between the tips of 

the fingers and a glass plate such that longitudinal and transverse cracks appear at a rolled diameter 

of 3 mm. At this point the soil has a stiff consistency. 

The liquid limit of a soil can be determined using the cone penetrometer or the Casagrande 

apparatus (BS 1377:1990:part 2, clauses 4.3, 4.5 / ASTM D-423-54T and ASTM D-424-54T). One of 

the major changes introduced by the 1975 British Standard (BS 1377 ) was that the preferred 

method of liquid limit testing became the cone penetrometer. This preference is reinforced in the 

revised 1990 British Standard which refers to the cone penetrometer as the ‘definitive method’. The 

cone penetrometer is considered a more satisfactory method than the alternative because it is 

essentially a static test which relies on the shear strength of the soil, whereas the alternative 

Casagrande cup method introduces dynamic effects. In the penetrometer test, the liquid limit of the 

soil is the moisture content at which an 80 g, 300 cone sinks exactly 20 mm into a cup of remoulded 

soil in a 5s period. 

Plasticity tests are widely used for classification of soils (Fig. 4.2) into groups on the basis of their 

position on the Casagrande chart (Casagrande 1948), but in addition they are used to determine the 

suitability of wet cohesive fill for use in earthworks, and to determine the thickness of sub-base 

required beneath highway pavements (Road Research Laboratory 1970). 

4-6 March 2009


FOR SOILS 

Plasticity index (Liquid limit – plastic limit) (%) 

Liquid 

limit (%) 

Figure 

4.2 Casagrande Plot Showing Classification of Soil into Groups 

4.6.2 

Chemical and Electro-chemical Tests 

During site investigation it is often necessary to 

carry out laboratory testss to determine the effects of 

the sub-soiused to check the soundness of aggregates for concrete or 

soil cement, 

to determine if electrolytic 

corrosion 

of metals will take place, 

or simply to act as index tests. 

or groundwater on concrete to be 

placed as foundations. Chemical tests may also be 

The effects of aggressive ground are numerous. Details can 

be found in 

Neville (1977), BRE Digest 

250 (1981), Tomlinson (1980) and 

BS 5930:1981. The available tests include those listed in Table 

4.3. 

Table 4.3 Available Chemical Tests 

Test 

Organic matter content 

Loss on ignition or ash content 

Sulphate content of soil and groundwater 

Carbonate content 

Chloride content 

Total dissolve solids 

pH value 

Resistivity 

Redox potential 

Source 

BS 

1377:part 3:1990, clause 

3 

BS 


4 

BS 


5 

BS 


6 

BS 


7 

BS 


8 

BS 


9 

BS 1377:part 3: 1990, clause 10 

BS 1377:part 3: 1990, clause 11 

The risk of acid attack should be 

assessed from pH data, depth to water table, the likelihood of 

water movement, the 

thickness of concrete, and whether 

it is subject to any hydrostatic head. 

Examples 

of low and high risk conditions are given below. 

March 2009 

4-7


FOR SOILS 

a. Low risk. pH 5.5— —7.0, stiff unfissured clay soil with water table below 

foundation level. 

b. Highh risk. pH < 3.5, permeable soil with water table above foundation level and risk 

groundwater movement. 

of 

Organic contents are also of use in classifying 

organic soilss such as peats. For most purposes the 

determination of ‘losss on ignition’ or ash conten 

is sufficient, but it should be remembered that this 

method tends to yield 

organic contents which may be up to 15% too high because the oven-dried 

specimen 

is fired at about 800—900°C and clay 

minerals and 

carbonates are altered. 

4.6.3 

Compaction Related Tests 

British Standard BS 1377: 1990:part 4 provides three specifications for laboratory compaction: 

a. 2.5 kg rammer method; 

b. 4.5 kg rammer method; and 

c. Vibrating hammer 

method for granular soils. 

Laboratory compaction tests are intended to model the field process, 

and to indicate the most 

suitable moisture content for compaction (the ‘optimum moisture content’) at whichh the maximum 

dry density will be achieved for a particular soil. 

The 2.5 kg rammer method is derivedd from the work 

of Proctor (1933) which introduced a test intended to be relevant to the compaction 

techniques in 

use in earthfill dam construction in 

the USA in the 1930s. The test subsequently became adopted by 

the American Association of State 

Highway Officials (AASHO), and was known as the Proctor or 

AASHO compaction test. 

A typical compaction curve (i.e. dry 

density as a function of moisture content) is shown 

in Fig. 4.3. 

Dry density (Mg/m 3 ) 

Moisture content (%) 

Figure 4.3 Typical Compaction Curves 

4-8 

March 2009


4.6.4 Compressibility, Permeability and Durability Tests 

Laboratory determinations of the permeability of granular soils can be made using the constant head 

and falling head permeameter tests (BS 1377: part 5:1990, clause 5). For granular soils any values 

of permeability must be regarded as approximate, since several important factors affect the accuracy 

of these tests. 

Cohesive soils can be tested for coefficient of permeability in the laboratory, and indeed it was for 

this purpose that Terzaghi (1923) produced the one-dimensional consolidation theory. Terzaghi 

noted that smear on the specimen boundaries greatly affected the measured soil permeability in his 

permeameter tests, and used an oedometer test in order that all water flow would occur out of the 

sample. Thus the coefficient of permeability can be obtained from triaxial or hydraulic consolidation 

tests since: 

k = c v m v w (4.4) 

where k = coefficient of permeability, c v = coefficient of consolidation, m v = the coefficient of 

compressibility, and w = density of water. 

4.6.5 Consolidation and Permeability Tests in Hydraulic Cells and with Pore 

Pressure Measurement 

Consolidation tests are frequently required either to assess the amount of volume change to be 

expected of a soil under load, for example beneath a foundation, or to allow prediction of the time 

that consolidation will take. The effect of predictions based on consolidation test results can be very 

serious, for example leading to the use of piling beneath structures, and the use of sand drains or 

stage construction for embankments. It is therefore important to appreciate the limitations of the 

commonly available test techniques. Three pieces of apparatus are in common use for consolidation 

testing. These are: 

a. The oedometer (Terzaghi 1923; Casagrande 1936); 

b. The triaxial apparatus (Bishop and Henkel 1962); and 

c. The hydraulic consolidation cell (Rowe and Barden 1966). 

a. Casagrande oedometer test 

The Casagrande oedometer test is most widely used. BS 1377: part 5:1990, clause 3 gives a 

standard procedure for the test. In this procedure the specimen is subjected to a series of preselected 

vertical stresses (e.g. 6, 12, 25, 50, 100, 200, 400, 800, 1600, 3200 kN/m2) each of which 

is held constant while dial gauge measurements of vertical deformation of the top of the specimen 

are made, and until movements cease (normally 24 h). 

b. Triaxial Dissipation Test 

The measurement of consolidation characteristics can be carried out in the triaxial dissipation test 

(Fig 4.6). The most common size of specimen is 102mm high x 102mm dia., and the test is carried 

out in a triaxial chamber such as might be used for a consolidated undrained triaxial compression 

test with pore pressure measurement. The specimen is compressed under the isotropic effective 

stress produced by the difference between the cell pressure and the back pressure, and volume 

change is recorded as a function of time, as in the consolidation stage of an effective strength triaxial 

compression test, but in addition pore pressure is measured at the base of the specimen. Drainage 

occurs upwards in the vertical direction but soil compression is three-dimensional, and for this reason 

the results of this test are not strictly comparable with those of an oedometer test. The 

compressibility determined from volume changes during the triaxial dissipation is greater than that 

March 2009 4-9


measured under conditions of zero lateral strain, and the difference is most pronounced for 

overconsolidated clays and compacted soils. 

c. Hydraulic Consolidation Cell (Rowe Cells Consolidation Test) 

The conventional oedometer enables one to determine the consolidation characteristics in the 

vertical direction only. With some modifications, the hydraulic consolidation cell (Rowe cell) with 

radial drainage can measure the horizontal consolidation properties. The Rowe cell is an incremental 

loading test similar to a conventional oedometer test with a reasonably long testing duration. These 

cells, in which load is applied to the sample hydraulically, offer many advantages and considerably 

widen the scope of laboratory testing. In addition, the hydraulic loading system gives accurate 

control of applied loads over a wide range, including high pressures on large diameter samples. 

(a) Schematic Diagram of Oedometer 

(b) Hydraulic Consolidation Cell 

Figure 4.4 Consolidation Test Apparatus 

4.6.6 Shear Strength Tests (Total and Effective Stresses) 

The principal tools available for strength determination include the California Bearing Ratio (CBR) 

apparatus, the Franklin Point Load Test apparatus (Franklin et al. 1971; Broch and Franklin 1972), 

the laboratory vane apparatus and various forms of direct shear and triaxial apparatus. For the 

purpose of relevance and application to DID related works, only the vane apparatus and the direct 

shear and triaxial tests are presented herein. 

Laboratory vane test 

The principles involved in the vane test are discussed in Section 3.3. Whilst the field vane typically 

uses a blade with a height of about 150 mm, the laboratory vane is a small-scale device with a blade 

height and width of about 12.7mm. The small size of the laboratory vane makes the device 

unsuitable for testing samples with fissuring or fabric, and therefore it is not very frequently used. 

The laboratory vane test is described in BS 1377: part 7:1990, clause 3. 

Direct shear test 

The vane apparatus induces shear along a more or less predetermined shear surface. In this respect 

the direct shear test carried out in the shear box apparatus (Skempton and Bishop 1950) is similar. 

Fig. 4.5 shows the basic components of the direct shear apparatus; soil is cut to fit tightly into a box 

which may be rectangular or circular in plan (Akroyd 1964; Vickers 1978; ASTM Part 19; Head 1982; 

BS 1377:1990), and is normally rectangular in elevation. The box is constructed to allow 

displacement along its horizontal mid-plane, and the upper surface of the soil is confined by a 

loading platen through which normal stress may be applied. Shear load is applied to the lower half of 

4-10 March 2009


the box, the upper half being restrained by a proving ring or load cell which is used to record the 

shear load. The sample is not sealed in the shear box; it is free to drain from its top and bottom 

surfaces at all times. 

The cross-sectional area over which the specimen is sheared is assumed to remain constant during 

the test. 

The direct shear test has been used to carry out undrained and drained shear tests, and to 

determine residual strength parameters. Morgenstern and Tchalenko (1967) reported the results of 

optical measurements on clays at various stages during the direct shear test, and it is clear that at 

peak shear stress and beyond, failure structures (Reidels and thrust structures) are not coincident 

with the supposed imposed horizontal plane of failure. In addition, the restraints of the ends of the 

box create an even more markedly non-uniform shear surface. Since the direction of the failure 

planes, the magnitude and directions of principal stresses and the pore pressure are not 

determinable in a normal shear box experiment, its results are open to various interpretations (Hill 

1950), and this test is now rarely used to determine undrained or peak effective strength 

parameters. Triaxial tests may be performed more conveniently and with better control. 

March 2009 4-11


FOR SOILS 

Figure 4.5 Bishop Direct Shear Box 

Triaxial Test 

The triaxial apparatus has been described in great detail by Bishop and Henkel (1962). The test 

specimen 

is normally a cylinder with an aspect ratio of two, 

which is sealed on its sides by a rubber 

membrane attached by rubber ‘O’ 

rings to a base pedestal 

and top cap 

(Fig. 4.6). Water pressure 

inside the 

cell provides the horizontal principal total stresses, 

while the vertical pressure at the top 

4-12 

March 2009


cap is produced by the cell fluid pressure and the ram force. The use of an aspect ratio of two 

ensures that the effects of the radial shear stresses between soil, and top cap and base-pedestal are 

insignificant at the centre of the specimen. 

The triaxial apparatus requires one or two self-compensating constant pressure systems, a volume 

change measuring device and several water pressure sensing devices. The ram force may be 

measured outside the cell using a proving ring, but most modern systems now use an internal 

electrical load cell mounted on the bottom of the ram. The ram is driven into the triaxial cell by an 

electrical loading frame which will typically have a capacity of 5000 or 10000 kgf and is capable of 

running at a wide range of constant speeds; triaxial tests are normally carried out at a controlled rate 

of strain increase. 

The three most common forms of test are: 

Figure 4.6 Triaxial Cell 

a. The unconsoldiated undrained triaxial compression test, without pore water pressure 

measurement (BS 1377:part 7:1990. clause 8); 

b. The consolidated undrained triaxial compression test, with pore water pressure measurement 

(BS 1377:part 8:1990, clause 7); and 

c. The consolidated drained triaxial compression test, with volume change measurement (BS 

1377:part 8:1990, clause 8). 

The unconsolidated undrained triaxial compression test is carried out on ‘undisturbed’ samples of 

clay in order to determine the undrained shear strength of the deposit in situ. Pore pressures are not 

measured during this test and therefore the results can only be interpreted in terms of total stress. 

March 2009 4-13


Peak effective strength parameters (c' and φ') may be determined either from the results of 

consolidated undrained triaxial compression tests with pore pressure measurement or from 

consolidated drained triaxial compression tests. The former test is normally preferred because it can 

be performed more quickly and therefore more economically. 

The consolidated undrained triaxial compression test is normally performed in several stages, 

involving the successive saturation, consolidation and shearing of each of three specimens. 

Saturation is carried out in order to ensure that the pore fluid in the specimen does not contain free 

air. If this occurs, the pore air pressure and pore water pressure will differ owing to surface tension 

effects: the average pore pressure cannot be found as it will not be known whether the measured 

pore pressure is due to the pore air or pore water, and at what level between the two the average 

pressure lies. 

The consolidation stage of an effective stress triaxial test is carried out for two reasons. First, three 

specimens are tested and consolidated at three different effective pressures, in order to give 

specimens of different strengths which will produce widely spaced effective stress Mohr circles. 

Secondly, the results of consolidation are used to determine the minimum time to failure in the shear 

stage. The effective consolidation pressures (i.e. cell pressure minus back pressure) will normally be 

increased by a factor of two between each specimen, with the middle pressure approximating to the 

vertical effective stress in the ground. 

Effective stress triaxial tests are far less affected by sample size effects than undrained triaxial tests, 

but the problems of sampling in stoney soils still make multistage testing an attractive proposition 

under certain circumstances. The effectiveness of this technique in consolidated undrained triaxial 

testing has been reported by Kenney and Watson (1961), Parry (1968) and Parry and Nadarajah 

(1973). 

The consolidated drained triaxial compression test, with volume change measurement during shear is 

carried out in a similar sequence to the consolidated undrained test, but during shear the back 

pressure remains connected to the specimen which is loaded sufficiently slowly to avoid the 

development of excess pore pressures. The coefficient of consolidation of the soil is derived in the 

manner described above from the volume change measurements made during the consolidation 

stage. 

Thus the shear stage of a drained triaxial test can be expected to take between 7 and 15 times 

longer than that of an undrained test with pore pressure measurement. 100mm dia. specimens of 

clay may require to be sheared for as much as one month. Once shearing is complete, the results 

are presented as graphs of principal stress difference and volume change as a function of strain, and 

the failure Mohr circles are plotted to give the drained failure envelope defined by the parameters cd' 

and φd' 

The effective strength parameters defined by drained triaxial testing should not be expected to be 

precisely the same as those for an undrained test, since volume changes occurring at failure involve 

work being done by or against the cell pressure (Skempton and Bishop 1954). In practice the 

resulting angles of friction for cohesive soils are normally within 1—2°, and the cohesion intercepts 

are within 5 kN/m 2 . The results of tests on sands can vary very greatly (for example, Skinner 1969). 

Stiffness tests 

From the 1950s through to the early 1980s there has been a preoccupation in commercial soil testing 

with the measurement of strength with less emphasis being paid to the measurement of detailed 

stress—strain properties such as stiffness. This is reflected in both the 1975 and the 1990 editions of 

BS 1377, both of which fail to consider the measurement of stiffness. 

4-14 March 2009


In most soils any discontinuities such as fissures will generally have a stiffness that is similar to that 

of the intact soil such that the intact soil stiffness may be used to predict with reasonable accuracy 

ground deformations and stress distributions. This means that laboratory triaxial tests on good 

quality ‘undisturbed’ specimens may yield adequate stiffness parameters for design purposes. 

However, conventional measurements of axial deformation of triaxial specimens, made outside the 

triaxial cell, introduce significant errors in the computation of strains. 

March 2009 4-15


REFERENCES 

[1] American Association of State Highway and Transportation Officials (AASHTO). (1995). 

Standard specifications for transportation materials and methods of sampling and testing: part II: 

tests, Sixteenth Edition, Washington, D.C. 

[2] American Society for Testing & Materials. (2000). ASTM Book of Standards, Vol. 4, Section 

08 and 09, Construction Materials: Soils & Rocks, Philadelphia, PA. 

[3] Bishop, A. W., and Henkel, D. J. (1962). The Measurement of Soil Properties in the Triaxial 

Test, Second Edition, Edward Arnold Publishers, Ltd., London, U.K., 227 p. 

[4] Bishop, A. W., and Bjerrum, L. (1960). “The relevance of the triaxial test to the solution of 

stability problems.” Proceedings, Research Conference on Shear Strength of Cohesive Soils, 

Boulder/CO, ASCE, 437-501. 

[5] Bishop, A. W., Alpan, I., Blight, G.E., and Donald, I.B. (1960). “Factors controlling the 

strength of partially saturated cohesive soils.”, Proceedings, Research Conference on Shear Strength 

of Cohesive Soils, Boulder/CO, ASCE, 503-532. 

[6] Clarke, B.G. (1995). Pressuremeters in Geotechnical Design. International Thomson 

Publishing/UK, and BiTech Publishers, Vancouver. 

[7] Deere, D. U., and Miller, R. P. (1966). Engineering classification and index properties of 

intact rock, Tech. Report. No. AFWL-TR-65-116, USAF Weapons Lab., Kirtland Air Force Base, NM. 

[8] Gibson, R. E. (1953). "Experimental determination of the true cohesion and true angle of 

internal friction in clays." Proceedings, 3rd International Conference on Soil Mechanics and 

Foundation Engineering, Zurich, Switzerland, 126-130. 

[9] International Society for Rock Mechanics Commission (1979). “Suggested Methods for 

Determining Water Content, Porosity, Density, Absorption and Related Properties.” International 

Journal Rock Mechanics. Mining Sci. and Geomechanics Abstr., Vol. 16, Great Britian, 141-156. 

[10] Jamiolkowski, M., Ladd, C. C., Germaine, J. T., and Lancellotta, R. (1985). “New 

developments in field and laboratory testing of soils.” Proceedings, 11th International Conference on 

Soil Mechanics & Foundation Engineering, Vol. 1, San Francisco, 57-153. 

[11] Littlechild, B.D., Hill, S.J., Statham, I., Plumbridge, G.D. and Lee, S.C. (2000). 

“Determination of rock mass modulus for foundation design”, Innovations & Applications in 

Geotechnical Site Characterization (GSP 97), ASCE, Reston, Virginia, 213-228. 

[12] LoPresti, D.C.F., Pallara, O., Lancellotta, R., Armandi, M., and Maniscalco, R. (1993). 

“Monotonic and cyclic loading behavior of two sands at small strains”. ASTM Geotechnical Testing 

Journal, Vol. 16 (4), 409-424. 

[13] LoPresti, D.C.F., Pallara, O., and Puci, I. (1995). “A modified commercial triaxial testing 

system for small strain measurements”. ASTM Geotechnical Testing Journal, Vol. 18 (1), 15-31. 

[14] Poulos, S.J. (1988). “Compaction control and the index unit weight”. ASTM Geotechnical 

Testing Journal, Vol. 11, No. 2, 100-108. 

4-16 March 2009


[15] Richart, F. E. Jr. (1977). "Dynamic stress-strain relations for soils - State of the art report." 

Proceedings, 9th International Conference on Soil Mechanics and Foundation Engineering, Tokyo, 

605-612. 

[16] Tatsuoka, F. and Shibuya, S. (1992). “ Deformation characteristics of soils & rocks from field 

& lab tests.” Report of the Institute of Industrial Science 37 (1), Serial No. 235, University of Tokyo, 

136 p. 

[17] Tatsuoka, F., Jardine, R.J., LoPresti, D.C.F., DiBenedetto, H., and Kodaka, T. (1997). “Theme 

Lecture: Characterizing the pre-failure deformation properties of geomaterials”. Proceeedings, 14 th 

International Conf. on Soil Mechanics & Foundation Engineering, Vol. 4, Hamburg, 2129-2164. 

[18] Tavenas, F., LeBlond, P., Jean, P., and Leroueil, S. (1983). “The permeability of natural soft 

clays: Parts I and II”, Canadian Geotechnical Journal, Vol. 20 (4), 629-660. 





[21] Woods, R. D. (1978). "Measurement of soil properties - state of the art report." Proceedings, 

Earthquake Engineering and Soil Dynamics, Vol. I, ASCE, Pasadena, CA, 91-178. 

[22] Woods, R.D. (1994). "Laboratory measurement of dynamic soil properties". Dynamic 

Geotechnical Testing II (STP 1213), ASTM, West Conshohocken, PA, 165-190. 

[23] Wroth, C. P., and Wood, D. M. (1978). "The correlation of index properties with some basic 

engineering properties of soils." Canadian Geotechnical Journal, Vol. 15 (2), 137-145. 

[24] Wroth, C. P. (1984). "The interpretation of in-situ soil tests." 24th Rankine Lecture, 

Géotechnique, Vol. 34 (4), 449-489. 

[25] oud, T.L. (1973). “Factors controlling maximum and minimum densities of sands”. Evaluation 

of Relative Density, STP 523, ASTM, West Conshohocken/PA, 98-112. 

March 2009 4-17



4-18 March 2009

CHAPTER 5 INTERPRETATION OF SOIL PROPERTIES

Chapter 5 INTERPRETATION OF SOIL PROPERTIES 





5.1 INTRODUCTION .......................................................................................................... 5-1 

5.1.1 Reporting of Test Results ............................................................................... 5-1 

5.2 COMPOSITION AND CLASSIFICATION ........................................................................... 5-2 

5.2.1 Soil Classification and Geo-Stratigraphy ........................................................... 5-2 

5.2.2 Soil Classification by Soil Sampling and Drilling ................................................ 5-2 

5.2.3 Soil Classification by Cone Penetration Testing ................................................. 5-3 

5.3 DENSITY ..................................................................................................................... 5-5 

5.3.1 Unit Weight .................................................................................................. 5-5 

5.3.2 Relative Density Correlations .......................................................................... 5-7 

5.4 STRENGTH AND STRESS HISTORY ............................................................................... 5-11 

5.4.1 Drained Friction Angle of Sands ..................................................................... 5-11 

7.4.2 Pre-consolidation Stress of Clays ................................................................... 5-13 

5.4.3 Undrained Strength of Clays and Silts ............................................................ 5-17 

5.4.4 Lateral Stress State ...................................................................................... 5-20 

5.5 FLOW PROPERTIES .................................................................................................... 5-21 

REFERENCES ....................................................................................................................... 5-23 

March 2009 5-i


List of Table 


5.1 Representative Permeability Values for Soils 5-22 



5.1 Delineation of Geostratigraphy and Soil & Rock Types by Drill & Sampling 

Methods 5-3 

5.2 Factors Affecting Cone Penetrometer Test Measurements in Soils (Hegazy, 1998) 5-4 

5.3 Chart for Soil Behavioral Classification by CPT (Robertson, Et Al., 1986) 5-5 

5.4 Interrelationship between Saturated Unit Weight and In-Place Water Content Of 

Geo-Materials 5-6 

5.5 Interrelationship between Minimum and Maximum Dry Densities of Quartz Sands 5-8 

5.6 Maximum Dry Density Relationship with Sand Uniformity Coefficient 5-9 

5.7 Relative Density Of Clean Sands From Standard Penetration Test Data 5-10 

5.8 Relative Density Evaluations Of NC and OC Clean Quartz Sands from CPT Data 5-11 

5.9 Typical Values of ø’ and Unit Weight for Cohesionless Soils 5-12 

5.10 Peak Friction Angle Of Sands From SPT Resistance 5-12 

5.11 Peak Friction Angle Of Un-Aged Clean Quartz Sands From Normalized CPT Tip 

Resistance 5-13 

5.12 Representative Consolidation Test Results in Overconsolidated Clay 5-14 

5.13 Trends for Compression and Swelling Indices in Terms of Plasticity Index 5-15 

5.14 Ratio Of Measured Vane Strength To Preconsolidation Stress (Suv/P') Vs. Plasticity 

Index (Ip) (After Leroueil And Jamiolkowski. 1991) 5-15 

5.15 Pre-consolidation Stress Relationship with Net Cone Tip Resistance from Electrical 

CPT 5-16 

5.16 Relationship Between Pre-consolidation Stress and Excess Porewater Pressures from 

Piezocones 5-16 

5.17 Relationship Between Pre-consolidation Stress and DMT Effective Contact Pressure in 

Clays 5-17 

5.18 Relationship between Preconsolidation Stress and Shear Wave Velocity in Clays 5-17 

5.19 Normalized Undrained Strengths for NC Clay under Different Loading Modes by 

Constitutive Model (Ohta, et al., 1985) 5-19 

5.20 Undrained Strength Ratio Relationship with OCR and ' for Simple Shear Mode 5-20 

5-ii March 2009



5 INTERPRETATION OF SOIL PROPERTIES 

The results of the field and laboratory testing program must be compiled into a simplified 

representation of the subsurface conditions that includes the geo-stratigraphy and interpreted 

engineering parameters. Natural geo-materials are particularly difficult to quantify because they 

exhibit complex behavior and involve the actions and interactions of literally infinite numbers of 

particles that comprise the soil and/or rock mass. In contrast to the more "well-behaved" civil 

engineering materials, soils are affected by their initial stress state, direction of loading, composition, 

drainage conditions, and loading rate. Thus, the properties of soil and rock properties must be 

evaluated through a program of limited testing and sampling. In certain cases, the soil properties 

may be altered or changed using ground modification techniques. 

All interpretations of geotechnical data will involve a degree of uncertainty because of the differing 

origins, inherent variability, and innumerable complexities associated with natural materials. The 

interpretations of soil parameters and properties will rely on a combination of direct assessment by 

laboratory testing of recovered undisturbed samples and in-situ field data that are evaluated by 

theoretical, analytical, statistical, and empirical relationships. 

The application of empirical correlations and theoretical relationships should be done carefully, with 

due calibration and verification with the companion sets of laboratory tests, to ensure that proper 

site characterization is achieved. Notably, many interrelationships between engineering properties 

and field tests have developed separately from individual sources, with different underlying 

assumptions, reference basis, and specific intended backgrounds, often for a specific soil. 

5.1.1 Reporting of Test Results 

Reporting of test results (field and laboratory) are presented in two basic forms. 

a. Factual Report 

b. Interpretative Report 

Factual Reports is a compilation of all the location plan of boreholes and test pits, borelogs, test pit 

logs, test results (field and laboratory) and photographs of site investigation activities without 

detailed interpretation of the test results. This report is basically presented by the S.I Contractor for 

their Client. 

Interpretative reports include the Factual Report as well as an interpretation of the test results by a 

geotechnical engineer/ expert to be used by the designers. This report can also be prepared by the 

S.I contractor by employing the services of a geotechnical engineer or it is prepared separately by 

the Client employing a geotechnical engineer depending on the nature of the site investigation 

contract. The interpretative report presents the interpretation of soil properties from in-situ tests 

and laboratory test for the analysis and design of foundations, embankments, slopes, and earthretaining 

structures in soils. Correlation of properties to laboratory index tests and typical ranges of 

values are also provided to check the reasonableness of field and laboratory test results. Reference is 

made to relevant established documents and standards in order to familiarize with appropriate and 

more detailed directions on the procedures and methodologies, as well as examples of data 

processing and evaluation. 

March 2009 5-1


5.2 COMPOSITION AND CLASSIFICATION 

Soil composition includes the relative size distributions of the grain particles, their constituent 

characteristics (mineralogy, angularity, shape), and porosity (density and void ratio). These can be 

readily determined by the traditional approach to soil investigation using a drilling and sampling 

program, followed by laboratory testing. 

The behavior of soil materials is controlled not only by their constituents, but also by less tangible 

and less quantifiable factors as age, cementation, fabric (packing arrangements, inherent structure), 

stress-state anisotropy, and sensitivity. In-situ tests provide an opportunity to observe the soil 

materials with all their relevant characteristics under controlled loading conditions. 

5.2.1 Soil Classification and Geo-Stratigraphy 

In the field, there are three approaches to soil classification and the delineation of geo-stratigraphy, 

i.e., drilling and sampling, cone penetration, and flat plate dilatometer soundings. 

Testing by the cone and dilatometer, measure the in-situ response of soil while in its original position 

and environment, thus indicating a "soil behavioural" type of classification at the moment of testing. 

The field tests are primarily conducted by deployment of vertical soundings to determine the type, 

thickness, and variability of soil layers, depth of bedrock, level of groundwater and presence of 

lenses, seams, inclusions, and/or voids. 

5.2.2 Soil Classification by Soil Sampling and Drilling 

Routine samplings involve the recovery of auger cuttings, drive samples, and pushed tubes from 

rotary-drilled boreholes. The boring may be created using solid flight augers (depth, z < 10 m), 

hollow-stem angers (z < 30 m), wash-boring techniques (z < 90 m), and wire-line techniques 

(applicable to 200 m or more). At select depths, split-barrel samples are obtained and a visualmanual 

examination of the recovered samples is sufficient for a general quantification of soil type. 

These 0.3-m long drive samples are collected only at regular 1.5-m intervals, however, and thus 

reflect only a portion of the subsurface stratigraphy. Less frequently, thin-walled undisturbed tube 

samples are obtained. More recently, sampling by a combination of direct-push and percussive forces 

has become available (e.g., geoprobe sampling; sonic drilling), whereby 25-mm diameter 

continuously-lined plastic tubes of soil are recovered. Although disturbed, the full stratigraphic profile 

can be examined for soil types, layers, seams, lenses, color changes, and other details. 

5-2 March 2009

Chapter 


Figure 5.1 Delineation of Geostratigraphy and Soil & Rock Types by Drill & Sampling Methods 

5.2.3 

Soil Classification by Cone Penetration Testing 

The cone penetrometer provides indirect assessments of soil classification type (in the classical 

sense) by measuring the responsee during full-displacement. of tip resistance (q c ), sleeve friction (f s ), and porewater 

pressures (u b ) are affected by the particle sizes, mineralogy, soil fabric, age, stress state, and other 

factors, as depicted in 

Fig. 5.2 (Hegazy, During a cone penetration test (CPT), 

the continuously recorded measurements 

1998). 

March 2009 

5-3

Chapter 


Figure 5.2 Factors Affecting Cone Penetrometer Test Measurements 

in Soils (Hegazy, 1998) 

A general rule of thumb is that the tip stress in sands is q t > 40 atm (Note: one atmosphere ≈ 1 

kg/cm 2 ≈ 1 tsf ≈ 100 kPa), while in many soft 

to stiff clays 

and silts, q t < 20 atm. In clean sands, 

penetration porewater pressures are near hydrostatic values 

(u 2 ≈ u o ≈ γ w w.z) since the permeability is 

high, while in soft to stiff intact clays, measured u 2 are often 3 to l0 times u o . Notably, in fissured 

clays and silts, the shoulder porewater readings can be 

zero or negative (up to minus one 

atmosphere, or - 100 kPa). With the sleeve friction reading (f s ), a processed value termed the friction 

ratio (FR) used: 

CPT Friction Ration, FR = R f = f s /q t 

(5.1) 

With CPT 

data, soil classification can be accomplished using a combination of two readings (either 

and f s or q t and u o ), 

or with all three readings. For this, it is convenient to definee a normalised 

porewater pressure parameter, B q , defined by: 

q t 

5-4 

March 2009

Chapter 


Porewater Pressure Parameter, B q = 

- 

- 

(5.2) 

chart using q t , FR, and 

B q is presented in Fig. 5. .3, indicating twelve classification regions. 

Figure 5.3 Chart for Soil Behavioral Classification by 

CPT (Robertson, Et Al., 1986) 

5.3 

5.3.1 

DENSITY 

Unit Weight 

The calculations of overburden stresses within a soil mass require evaluations of the 

unit weight or 

mass density of the various strata. Unit weight is defined as soil weight per unit volume (units of 

kN/m 3 ) and denoted by the symbol . Soil mass density is measured as 

mass per volume (in either 

g/cc or kg/m 3 ) and denoted by . In common 

use, the terms "unit weight" and "density" are used 

interchangeably. Their interrelationship is: 

γ = ρ.g 

(5.3) 

where g = gravitational constantt = 9.8 m/sec 2 . A reference value for fresh water is adopted, 

whereby ρ w = 1 g/cc, and the corresponding 

γ w = 9.8 kN/ /m 3 . In the laboratory, soil unit weight is 

measured on tube samples of natural soils and depends upon the specific gravity of solids (Gs), 

water content (w n ), and void ratio (e o ), as well as the degreee of saturation (S). These parameters are 

interrelated by the soil identity: 

Gs w n = S e o 

(5.4) 

where S = 1 (100%) for saturated soil (generally assumed for soil layers lying below the 

groundwater table) and S = 0 (assumed for granular soils above the water table). 

March 2009 

5-5


For the case of clays and silts above the water table, the soils may have degrees of saturation 

between 0 to 100%. Full saturation can occur due to capillarity effects and varies as the atmospheric 

weather. The identity relationship for total unit weight is: 

γ T = 1+w n 

1+e o G sγ w 

The estimation of unit weights for dry to partially saturated soils depends on the degree of 

saturation, as defined by (5.4) and (5.5). 

Figure 5.4 Interrelationship between Saturated Unit Weight and In-Place Water Content Of Geo- 

Materials 

The total overburden stress (σ vo ) is calculated from: 

σ vo = ∑ T ∆z (5.6) 

which in turn is used to obtain the effective vertical overburden stress: 

σ vo ’ = σ vo - u o (5.7) 

where the hydrostatic porewater pressure (u o ) is determined from the water table. 

5-6 March 2009


5.3.2 Relative Density Correlations 

The relative density (D R ) is used to indicate the degree of packing of sand particles and applicable 

strictly to granular soils having less than l5 percent fines. The relative density is defined by: 

D R = e max-e o 

e max -e min 

(5.8) 

where e max = void ratio at the loosest state and e min = void ratio at the densest state. The direct 

determination of D R by the above definition is not common in practice, however, because three 

separate parameters (e o , e max , and e min ) must be evaluated. 

For a given soil, the maximum and minimum void states are apparently related (Poulos, 1988). A 

compiled database indicates (n = 304; r 2 = 0.851; S.E. = 0.044): 

e min = 0.571 e max (5.9) 

For dry states (w = 0), the dry density is given as: d = Gs. γ w /(l+e) and the relationship between 

the minimum and maximum densities is shown in Fig. 7.5 for a variety of sands. The mean trend is 

given by the regression line: 

d (min) = 0.808 d(max) (5.10) 

Laboratory studies by Youd (1973) showed that both e max and e min depend upon uniformity 

coefficient (UC = D 60 /D 10 ), as well as particle angularity. For a number of sands (total n = 574), this 

seems to be borne out by the trend presented in Fig. 5.6 for the densest state corresponding to e min 

and d (max) . The correlation for maximum dry density [ d (max) ] in terms of UC for various sands is 

shown in Fig 5.7 and expressed by (n = 574; r 2 = 0.730): 

d(max) = 9.8 [1.65 + 0.52 log (UC)] (5.11) 

March 2009 5-7


Figure 5.5 Interrelationship between Minimum and Maximum Dry Densities of Quartz Sands. 

(Note: Conversion in terms of mass density and unit weight = 1 g/cc = 9.8 kN/m 3 = 62.4 pcf) 

5-8 March 2009


Figure 5.6 Maximum Dry Density Relationship with Sand Uniformity Coefficient (UC = D60/D10). 

(Note: Conversion In Terms Of Mass Density And Unit Weight: 1 G/Cc = 9.8 Kn/M3 = 62.4 Pcf) 

From a more practical stance, in-situ penetration test data are used to evaluate the in-place relative 

density of sands. The original D R relationship for the SPT suggested by Terzaghi & Peck (1967) has 

been re-examined by Skempton (1986) and shown reasonable for many quartz sands. The 

evaluation of relative density (in percent) is given in terms of a normalized resistance [(N 1 ) 60 ], as 

shown in Fig. 5.7. 

D R = 100 N 1 60 

60 

(5.12) 

where (N 1 ) 60 = N 60 /(σ. vo’ ) 0.5 is the measured N-value corrected to an energy efficiency of 60%oand 

normalised to a stress level of one atmosphere. Note here that the effective overburden stress is 

given in atmospheres. In a more general fashion, the normalised SPT resistance can be defined by: 

(N 1 ) 60 = N 60 /(σ vo’ /p a ) 0.5 for any units of effective overburden stress, where p a is a reference stress = 

1 bar ≈ 1 kg/cm 2 ≈ 1 tsf ≈ 100 kPa. The range of normalized SPT values should be limited to (N 1 ) 60 < 

60, since above this value, apparent grain crushing occurs due to high dynamic compressive forces. 

Additional effects of over-consolidation, particle size, and aging may also be considered, as these too 

affect the correlation (Skempton, 1986; Kulhawy & Mayne, 1990). 

A comparable approach for the CPT can be made based on calibration chamber test data on clean 

quartz sands (Fig. 5.8). The trends for relative density (in percent) of unaged uncemented sands 

are: 

March 2009 5-9


Overconsolidated sands: D R = 100 

q n 

0.2 (5.13a) 

300 OCR 

Normally-consolidated Sands: D R = 100 q n 

300 

(5.13b) 

where q t1 = q c /(σ vo’ ) 0.5 is the normalized tip resistance with both the measured q c and the effective 

overburden stress are in atmospheric units. The relationship should be restricted to q t1 < 300 

because of possible grain crushing effects. For any units of effective overburden stress and cone tip 

resistance, the normalized value is given by: q t1 = (q t /p a )/(σ vo ‘ /p a ) 0.5 , where p a is a reference stress 

= l bar ≈ 1 kg/cm 2 ≈ 1 tsf ≈ l00kPa. 

Figure 5.7 Relative Density Of Clean Sands From Standard Penetration Test Data 

Note: Normalized Value (N 1 ) 60 = N 60 /(σ. Vo’ ) 0.5 Where σ Vo ’ is In Units Of Bars Or Tsf. 

5-10 March 2009


Figure 5.8 Relative Density Evaluations Of NC and OC Clean Quartz Sands from CPT Data. 

Note: Normalized resistance is q t1 = q c /(σ’ Vo ) 0.5 with stresses in atmospheres (1 Atm=1 Tsf=100 Kpa). 

5.4 STRENGTH AND STRESS HISTORY 

The results of in-situ test measurements are convenient for evaluating the strength of soils and their 

relative variability across a project site. For sands, the drained strength corresponding to the 

effective stress friction angle (ø') is interpreted from the SPT, CPT, DMT, and PMT. For short-term 

loading of clays and silts, the undrained shear strength (c u ) is appropriate and best determined from 

normalized relationships with the degree of over-consolidation. In this manner, in-situ test data in 

clays are used to evaluate the effective pre-consolidation stress (σ p ') from CPT, CPTu, DMT, and V s , 

which in turn provide the corresponding over-consolidation ratios (OCR = σ p '/σ vo '). 

The long-term strength of intact clays and silts is represented by the effective stress strength 

parameters (ø’ and c’ = 0) that are best determined from either consolidated undrained triaxial tests 

with pore water pressure measurements, drained trail tests, or slow direct shear box tests in the lab. 

For fissured clay materials, the residual strength parameters (o r ’ and c ry ’ = 0) may be appropriate, 

particularly in slopes and excavations, and these values should be obtained from either laboratory 

ring shear tests or repeated direct shear box test series. 

5.4.1 Drained Friction Angle of Sands 

The peak friction angle of sands (ø') depends on the mineralogy of the particles, level of effective 

confining stresses, and the packing arrangement (Bolton, 1986). Sands exhibit a nominal value of ø' 

due solely to mineralogical considerations that corresponds to the critical state (designated r ocs '). The 

critical state represents an equilibrium condition for the particles at a given void ratio and effective 

confining stress level. For clean quartzite sands, a characteristic r ocs ' ≈ 33 o , while a feldspathic sand 

may show ø cs ' ≈ 30 o and a micaceous sandy soil exhibit ø cs ' ≈ 27 o . Under many natural conditions, the 

sands are denser than their loosest states and dilatancy effects contribute to a peak ø' that is greater 

than ø cs '. Fig. 5.9 shows typical values of ø' and corresponding unit weights over the full range of 

cohesionless soils. 

March 2009 5-11


Figure 5.9 Typical Values of ø’ and Unit Weight for Cohesionless Soils. (NAVFAC DM 7.1, 1982) 

The effective stress friction angle (ø') of sand is commonly evaluated from in-situ test data. The peak 

friction angles (ø') in terms of the (N 1 ) 60 resistances are presented in Fig. 5.10. 

Figure 5.10 Peak Friction Angle Of Sands From SPT Resistance (Data From Hatanaka & Uchicla, 

1996). Note: The Normalised Resistance Is (N 1 ) 60 = N 60 /(σ Vo’ /P a ) 0.5 , Where P a = 1 Bar = 1 Tsf = 100 

Kpa 

5-12 March 2009


The cone penetrometer can be considered a miniature pile foundation and the measured tip stress 

(q T ) represented the actual end bearing resistance (q b ). In bearing capacity calculations, the pile end 

bearing is obtained from limit plasticity theory that indicates: q b = N q. σ vo ' where N q is a bearing 

capacity factor for surcharge and depends upon the friction angle. Thus, one popular method of 

interpreting CPT results in sand is to invert the expression (N q = q T /σ vo ') to obtain the value of φ' 

(e.g., Robertson & Campanella, 1983). One method for evaluating the peak φ’ of clean quartz sands 

from normalized CPT tip stresses is presented in Fig. 5.11. 

Figure 5.11 Peak Friction Angle Of Un-Aged Clean Quartz Sands From Normalized CPT Tip 

Resistance. (Calibration Chamber Data Compiled By Robertson & Campanella, 1983). 

7.4.2 Pre-consolidation Stress of Clays 

The effective preconsolidation stress σ p ', is an important parameter that governs the strength, 

stiffness, geostatic lateral stress state, and porewater pressure response of soils. It is best 

determined from one-dimensional oedometer tests (consolidation tests) on high-quality tube samples 

of the soil. Sampling disturbance, extrusion, and handling effects tend o reduce the magnitude of σ p ' 

from the actual in-place value. The normalised form is termed the overconsolidation ratio (OCR) and 

defined by: 

OCR = σ p ’/σ vo ’ (5.14) 

Soils are often over-consolidated to some degree because they are old in geologic time scales and 

have undergone many changes. Mechanisms causing over-consolidation include erosion, desiccation, 

groundwater fluctuations, aging, freeze-thaw cycles, wet-dry cycles, glaciation, and cementation. 

March 2009 5-13


A representative e-log(σ v ’) curve obtained from one-dimensional consolidation testing on a marine 

clay is presented in Fig. 5.12. The observed pre-consolidation stress is seen to separate the 

recompression phase ("elastic strains") from the virgin compression portion (primarily "plastic 

strains") of the response. 

A check on the reasonableness of the obtained compression indices may be afforded via empirical 

relationships with the plasticity characteristics of the clay. A long-standing expression for the 

compression index (C c ) in terms of the liquid limit (LL) is given by (Terzaghi, et al., 1996): 

C c = 0.009 (LL-10) (5.15) 

. 

In natural deposits, the measured C c may be greater than that given by (5.15) because of inherent 

fabric, structure, and sensitivity. For example, in the case in Fig. 5.12 with LL = 41, (5.15) gives a 

calculated C c = 0.33, vs. measured C c = 0.38 in the oedometer. 

Figure 5.12 Representative Consolidation Test Results in Overconsolidated Clay 

Statistical expressions for the virgin compression index (C c ) and the swelling index (C s ) from unloadreload 

cycles are given in Fig. 5.13 in relation to the plasticity index (PI). However, it should be 

noted that the PI is obtained on remoulded soil, while the consolidation indices are measurements on 

natural clays and silts. Thus, structured soils with moderate to high sensitivity and cementation will 

depart from these observed trends and signify that additional testing and care are warranted. 

5-14 March 2009

Chapter 


Figure 5.13 Trends for Compression and 

Swelling Indices in Terms of Plasticity 

Index 

In clays and silts, the profile of preconsolidation stress can be evaluated via in-situ test data. a 

relationship between p ', plasticity 

index (PI) and the (raw) measured vane strength (s uv ) is given in 

Fig. 5.14. This permits immediatee assessment 

of the degree of over-consolidation 

of natural soil 

deposits. 

Figure 

5.14 Ratio Of Measured Vane Strength 

To Preconsolidation Stress (Suv/P') Vs. Plasticity 

Index (Ip) (After Leroueil And Jamiolkowski. 1991) 

For the electric cone penetrometer, Fig. 5.15 shows a relationship for σ p ' in terms of net cone tip 

resistance (q T - σ vo ‘ ) for intact clay 

deposits. Fissured clays are seen to lie above this 

trend. For the 

piezocone, σ p ' can be evaluated from excess porewater pressures (u 1 - u o o), as seen in Fig. 5.16. 

March 2009 

5-15


A direct correlation between the effective pre-consolidation stress and effective contact pressure 

(p o -u o ) measured by the flat dilatometer is given in Fig. 5.17, again noting that intact clays and 

fissured clays respond differently. The shear wave velocity (V S ) can also provide estimates of σ p ', per 

Fig. 5.18. In all cases, profiles of σ p ' obtained by in-situ tests should be confirmed by discrete 

oedometer results. 

Figure 5.15 Pre-consolidation Stress Relationship with Net Cone Tip Resistance from Electrical CPT 

Figure 5.16 Relationship Between Pre-consolidation Stress and Excess Porewater Pressures from 

Piezocones 

5-16 March 2009


Figure 5.17 Relationship Between Pre-consolidation Stress and DMT Effective Contact Pressure in 

Clays 

Figure 5.18 Relationship between Preconsolidation Stress and Shear Wave Velocity in Clays. 

(Data from Mayne, Robertson, & Lunne, 1998) 

5.4.3 Undrained Strength of Clays and Silts 

The undrained shear strength (s u or c u ) is not a unique property of soils, but a behavioral response 

to loading that depends upon applied stress direction, boundary conditions, strain rate, overconsolidation, 

degree of fissuring, and other factors. Therefore, it is often a difficult task to directly 

compare undrained strengths measured by a variety of different 1ab and field tests, unless proper 

March 2009 5-17

Chapter 


accounting of these factors is giver due consideration and adjustmentss are made accordingly. For 

example, 

the undrained shear strength represents the failure condition corresponding 

to the peak of 

the shear stress vs. shear strain curve. The time to reach the peak is a rate effect, such that 

consolidated undrained triaxial tests are usually conductedd with a time-to-failure on the order of 

several hours, whereas a vane shear may take several minutes, yet in contrast to seconds by a cone 

penetrometer. 

For normally-consolidated clays and silts, Fig. 5.19 shows the relative hierarchy of these modes and 

the observed trends with plasticity 

index (I p ). In this presentation, the undrained shear strength has 

been normalized by the effective overburden stress level, as 

denoted by the ratio (s u / σ vo ', or c u /σ vo o'), 

that refers to the older c/p' ratio. 

Fig. 5.19 Modes of Undrained Shear Strength Ratio (s u /σσ vo ') NC for Normally-Consolidated Clays 

(Jamiolkowski, et al. (1985)). 

The theoretical interrelationships 

of undrainedd loading modes for normally consolidated clay are 

depicted in Fig. 5.20 using a constitutive model (Ohta, et al., 

1985). 

5-18 

March 2009


Figure 5.19 Normalized Undrained Strengths for NC Clay under Different Loading Modes by 

Constitutive Model (Ohta, et al., 1985) 

Based on extensive experimental data (Ladd, 1991) and critical state soil mechanics (Wroth, 1984), 

the ratio (s u /σ vo ') increases with over-consolidation ratio (OCR) according to: 

(s u /σ vo ’) OC = (s u /σ vo ’) NC OCR A (5.16) 

where A ≈ 1- C S /C C and generally taken to be about 0.8 for unstructured and uncemented soils. 

Thus, if a particular shearing mode is required, it can be assessed using either Figs. 5.19 or 5.20 to 

obtain the NC value and equation (5-16) to determine the undrained strength for over-consolidated 

states. In many situations involving embankment stability analyses and bearing capacity calculations, 

the simple shear mode may be considered an average and representative value of the undrained 

strength characteristics, as shown by Fig. 5.21 and given by: 

(s u /σ vo ’) DSS = ½ sin ’ OCR A (5.17) 

March 2009 5-19


Figure 5.20 Undrained Strength Ratio Relationship with OCR and ' for Simple Shear Mode 

5.4.4 Lateral Stress State 

The lateral geostatic state of stress (K o ) is one of the most elusive measurements in geotechnical 

engineering. It is often represented as the coefficient of horizontal stress K o = σ ho '/σ vo ' where σ ho ' = 

effective lateral stress and σ vo ' = effective vertical stress. A number of innovative devices have been 

devised to measure the in-place total horizontal stress (σ ho ) including: total stress cell (push-in 

spade), self-boring pressuremeter, hydraulic fracturing apparatus, and the Iowa stepped blade. 

Recent research efforts attempt to use sets of directionalised shear wave measurements to decipher 

the in-situ K o in soil formations. 

For practical use, it is common to relate the K o state to the degree of overconsolidation, such as: 

K 0 = (1 – sin ’) OCR sin ’ (5.18) 

which was developed on the basis of special laboratory tests including instrumented oedometer 

tests, triaxial cells, and split rings (Mayne & Kulhawy, 1982). Fig. 5.22 shows field data 

measurements of K o for clays and sands. 

5-20 March 2009


Fig. 5.22 Field K o - OCR Relationships for (a) Natural Clays and (b) Natural Sand 

In general, the value of K o has an upper bound value limited by the passive coefficient, K p . The 

simple Rankine value is given by: 

K p = tan 2 (45° + ½ ’) = (1 + sin ’)/(1 + sin ’) (5.19) 

When the in-situ K o reaches the passive value K p , fissures and cracks can develop within the soil 

mass. This can be important in sloped masses since extensive fissuring is often associated with 

drained strengths that are at or near the residual strength parameters (φ r ' and c r ' = 0). 

5.5 FLOW PROPERTIES 

Soils exhibit flow properties that control hydraulic conductivity (k), rates of consolidation, 

construction behaviour, and drainage characteristics in the ground. Field measurements for soil 

permeability have been discussed previously and include pumping tests with measured drawdown, 

slug tests, and packer methods. Laboratory methods are presented in Section 4.6.5 and include 

falling head and constant head types in permeameters. An indirect assessment of permeability can 

be made from consolidation test data. Typical permeability values for a range of different soil types 

are provided in Table 5.1. Results of pressure dissipation readings from piezocone and flat 

dilatometer and holding tests during pressuremeter testing can be used to determine permeability 

and the coefficient of consolidation (Jamiolkowski, et al. 1985). 

March 2009 5-21

Chapter 


Table 5.1 Representativ 

ve Permeability Values for Soils 

The permeability (k) can be determined from the dissipation test data, either by use of the direct 

correlative relationship presented earlier, or alternatively by the evaluation of the 

coefficient of 

consolidation c h . Assuming radial flow, the horizontal permeability (k h ) is obtained from: 

k h = 

(5.20) 

where D' 

= constrained modulus obtained from oedometer tests. 

5-22 

March 2009


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March 2009 

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Chapter 1 

Chapter 2 

Chapter 3 

Chapter 4 

Chapter 5 

Chapter 6 

GEOMATICS AND LAND SURVEY SERVICES 

MAP PROJECTION 

TYPES OF SURVEY 

REFERENCES ON GEOMATICS AND LAND SURVEY SERVICES 

GEOGRAPHICAL INFORMATION SYSTEM (GIS) 

CHECKLIST FOR TERRAIN FEATURES 

March 2009 

iii



iv March 2009

PART 3: ENGINEERING SURVEY

CHAPTER 1 GEOMATICS AND LAND SURVEY SERVICES

Chapter 1 GEOMATICS AND LAND SURVEY SERVICES 




1.1 OVERVIEW ............................................................................................................... 1-1 

1.2 FIELD OF GEOMATICS AND ENGINEERING SURVEYING ............................................... 1-1 

1.3 APPLICATION AREAS ................................................................................................. 1-1 

1.4 SOURCE OF MATERIAL FOR GEOMATIC PLANNING ...................................................... 1-1 

1.5 PRINCIPLES OF SURVEYING EXERCISED BY SURVEYORS ............................................. 1-2 

1.5.1 Basic Principles Adopted by Surveyors ........................................................... 1-2 

1.5.2 Control ........................................................................................................ 1-2 

1.5.3 Revision ................................................................................................................... 1-3 

1.5.4 Economy and Accuracy ................................................................................. 1-4 

1.5.5 The Independent Check ................................................................................ 1-4 

1.5.6 Safeguarding ............................................................................................... 1-4 

1.6 REFERENCES ............................................................................................................ 1-5 

March 2009 1-i




1.1 Types of Traverse 1-3 

1-ii March 2009


1.1 OVERVIEW 

1 GEOMATICS AND LAND SURVEY SERVICES 

Planning and proposals for, and later, implementation of Department of Irrigation and Drainage 

project of various types have to take into consideration survey information provided by geomatics 

and land survey services. Geomatics is a fairly new term. It includes the tools and techniques used 

in land surveying for engineering works, remote sensing, Geographic Information System (GIS), 

Global Positioning System (GPS) and related forms of earth mapping. Originally, used in Canada, the 

term geomatics has been adopted by the International Organization for Standardization, the Royal 

Institution of Chartered Surveyors, the Institution of Surveyors Malaysia and many other 

international authorities. Some, especially the United States, prefer to use the term geospatial 

technology. 

The rapid progress and increased utilization of geomatics has been made possible by advances in 

computer technology, computer science and software engineering as well as advances in remote 

sensing technologies which provide imagery using space borne and air borne sensors. 

1.2 FIELD OF GEOMATICS AND ENGINEERING SURVEYING 

a. Geodesy 

b. Surveying 

c. Mapping 

d. Positioning of structures 

e. Geomatic Engineering 

f. Navigation 

g. Remote Sensing 

h. Photogrammetry 

i. Geographic Information System 

j. Global Positioning System 

k. Geospatial Technology 

1.3 APPLICATION AREAS 

a. The environment 

b. Land management 

c. Urban planning 

d. Subdivision planning in land development and land acquisition 

e. Infrastructure management 

f. Natural and infrastructure resource monitoring 

g. Coastal erosion management and mapping 

h. Natural disaster information for disaster risk reduction and response 

1.4 SOURCE OF MATERIAL FOR GEOMATIC PLANNING 

a. In Malaysia the initial source for obtaining material and information to plan and then formulate 

the term of reference and scope of work for proposals can be obtained from:- 

b. Topographic maps and aerial photographs from the Mapping Division of the Department of 

Survey and Mapping [1] Department of Survey and Mapping Website: http://www.jupem.gov.my 

c. Cadastral Certified Plans and Cadastral Standard Sheets from the Cadastral Survey Division of 

the Department of Survey and Mapping 

d. Thematic or geological maps from the Mineral and Geosciences Department Natural Resources 

and Environment Ministry 

March 2009 1-1


e. Malaysian Centre for Geospatial Data Infrastructure (MaCGDI) Ministry of Natural Resources and 

Environment [2] MaCGDI Website:http:// www.mygeoportal.gov.my 

f. DigitalGlobe the provider of high resolution QuickBird Imagery. QuickBird’s high resolution 

satellite imagery is available with resolution of 1.6 ft or 50cm panchromatic to 2ft or 70cm 

panchromatic, natural colors, colors infrared or 4-band pan sharpened [3] Digital Globe Website: 

http:// www.digitalGlobe.com. Digital Globe images has to be obtained through the Malaysian 

Centre for Remote Sensing (MACRES) 

g. Combination in the supply of a mosaic assembled from Quick Bird Satellite Images supplied by 

Digital Globe and color aerial photographs supplied by the Department of Survey and Mapping 

Overlaid with Department of Survey and Mapping cadastral standard sheet information can be 

customized. e.g. Bertam area Kepala Batas 

h. US Army Corps of Engineer Hydrographic Manual EM1110-2-1003 from the Web. (Chapter 17 – 

River Engineering and Channel Stabilization Surveys). [4] US Army Corps of Engineers website 

available by keying in “us army corps of engineers hydrographic survey manual” then click 

“EM1110-2-1003” 

1.5 PRINCIPLES OF SURVEYING EXERCISED BY SURVEYORS 

1.5.1 Basic Principles Adopted by Surveyors 

Users are informed that regardless of changes in techniques and equipment, the basic principles of 

surveying, which have been tested and proved over the years by geomatics and land surveyors 

remain and are applicable to all types of surveying. They are:- 

a. Control comprising planimetric (Horizontal) and Height (Vertical) 

b. Revision 

c. Economy of Accuracy 

d. The independent check 

e. Save guarding 

1.5.2 Control 

Any survey, whether large or small, depends upon the establishment of a carefully measured control 

framework which contains measured points linked with lines which encompass the whole area to be 

surveyed. The measured lengths and bearings of these straight lines, known as traverses, linking 

these series of points to form the various types of traverses are shown in Fig 1.1 below. Subsequent 

work is then fitted inside this framework and is adjusted to it. All TBMs should be connected by a 

closed leveling net which contain height points linked by survey lines which is tied to a minimum of 2 

Survey Department Bench Marks (BM). Surveyors also check Azimuths or bearings reckoned from 

true north by solar observation of the sun at suitable intervals with maximum closing error of 

1:4,000 for traverses within the horizontal control network (as a guide only). 

An open traverse is not acceptable unless it is double checked, both by angles and distances. 

1-2 March 2009


KNOWN 

STATIONS 

A. CLOSED LOOP TRAVERSE 

KNOWN 

STATIONS 

KNOWN 

STATIONS 

B. CLOSED CONNECTION TRAVERSE 

KNOWN 

STATIONS 

C. OPEN TRAVERSE 

Figure 1.1 Types of Traverse 

1.5.3 Revision 

Whenever a survey is initiated, the methods and scope of works employed by the surveyor should be 

formulated in the light of the following requirements: 

a. The requirements of the team of professionals who will be designing and subsequently 

implementing the project for the Department of Irrigation and Drainage. Checks should also 

be made that the requirement of another Department is taken into consideration e.g. the 

Ministry of Agriculture, Public Works Department or the Land Office resettlement plan. 

b. It is important that a survey work done for one purpose may at some future date be used 

for a different purpose. The department concerned should anticipate this and consider 

whether, by some minor adjustment, the scope of works can be made more generally useful 

than the immediate needs. 

c. It is important that all leveling or height control and connection work which include the 

establishment of hydrological stations are tied to Survey and Mapping Department Bench 

Marks (BM) and that Temporary Bench Marks (TBM) are established on permanent features 

at strategic locations within the proposed scheme for future use. 

d. The field surveyor’s first task is to establish the horizontal and vertical control frameworks 

which are tied to the Survey Department Horizontal Datum for position and to the Land 

Survey Vertical Datum or the Chart Datum at the respective tide gauge stations for levels or 

TBM. Fitted within this framework are the supplementary control such as the DID proposed 

baseline, check line or secondary gridline where appropriate to pick up details of features 

and points contained in the Term of Reference (TOR) 

March 2009 1-3


1.5.4 Economy and Accuracy 

It is important, before any field survey operation is started, to weigh the accuracy against the time, 

resources and costs. The greater the accuracy required, the greater the cost of operation. Since 

accuracy depends upon the elimination or reduction of errors, it is essential that the surveyor 

understands the nature of the errors and plans his works in such a way to reduce them to acceptable 

levels to meet the misclosure tolerances adopted. 

1.5.5 The Independent Check 

In every survey operation it is the responsibility of a surveyor to do a check. It is best to employ a 

system which is completely self checking. Where this is not possible the check applied should be as 

independent as possible and not just a repetition of the previous operation. For example, if the 

measurement of the length is carried out, the check applied should be made by measuring the 

distance again using different unit of length or measuring in the reverse direction. In many cases a 

rough check is very useful and sometime all that is required. Computations which are not self 

checking should be completed by another survey staff including, using, if possible, methods other 

than those used. 

1.5.6 Safeguarding 

Marks established by the field surveyor for the horizontal and vertical control framework should be as 

permanent as possible or easily re-established from nearby marks. Liaison with Agricultural 

Department may be considered during planning for topographical surveys to coordinate simultaneous 

concurrent activities to collect water and soil test samples to determine their suitability for crop 

cultivation. Hydrological stations for systematic collection of data such as rainfall, stream flow, 

maximum flood levels, tidal range, etc. should also be considered. 

1-4 March 2009


REFERENCES 

[1] Department of Survey and Mapping website http://www.jupem.gov.my 

[2] Malaysian Centre for Geospatial Data Infrastructure (MaGDI) website 

http://www.mygeoportal.gov.my 

[3] Digital Globe for Satellite Imagery at website http://www.digitalglobe.com 

March 2009 1-5


(This page is intentionally blank) 

1-6 March 2009

CHAPTER 2 MAP PROJECTION

Chapter 2 MAP PROJECTION 




2.1 INTRODUCTION .......................................................................................................... 2-1 

2.2 Map Projection Malaysia .................................................................................. 2-2 

2.2.1 Rectified Skew Orthomorphic (RSO) Projection ..................................... 2-2 

2.2.2 Cassini Soldner Projection ................................................................................ 2-2 

2.2.3 WGS (World Geodetic System) 84 Ellipsoid ....................................................... 2-3 

2.2.4 GDM 2000 or Geocentric Datum Malaysia 2000 .................................................. 2-3 

2.3 REFERENCES ............................................................................................................... 2-5 

March 2009 2-i




2.1 The Ellipsoid 2-1 

2.2 RSO Grid Projection on Topographic Map 2-2 

2.3 Peninsular Malaysia GPS Network 2-4 

2-ii March 2009


2 MAP PROJECTION 


• A map projection is used to portray all or part of the round Earth by transforming/projecting it 

from a round surface (ellipsoid/spheroid) on to a plane or flat surface with some distortion. 

• Every projection has its own set of advantages and disadvantages. There is no “best” 

projection. 

• The mapmaker must select the one best suited to the needs, reducing distortion of the most 

important features. 

• Every flat map misrepresents the surface of the Earth in some way. No map can rival a globe in 

truly representing the surface of the entire Earth. However, a map or parts of a map 

constructed from map projections can show one or more but never all of the following. True 

directions or bearings. True distances or scale. True areas. True shapes. Hence mapmaking is 

an art and science of trade-offs. 

• Mapmakers and mathematicians have devised almost limitless equations to show the geographic 

image of the globe on paper. The mathematical model which is an approximation of the actual 

shape of the earth is commonly referred to as a spheroid or ellipsoid. 

North pole 

P 

Geoid 

b 

a 

Equatorial 

Plane 

P 1 

Ellipsoid 

Elements of an ellipse 

a = Semi Major Axis 

b = Semi Minor Axis 

f = Flattening = (a-b)/a 

PP’ = Axis of revolution of the earth's ellipsoid 

Figure 2.1 The Ellipsoid 

• As shown in the Figure 2.1 above the surface of the earth is not a sphere but an irregular 

changing shape, due to terrain features such as hills, mountains, valleys, rivers and the seas. 

This irregular surface has been approximated mathematically to that of an ELLIPSOID. 

Locations of topographic features on the curved surface of the ellipsoid earth are described in 

terms of latitude (Ø) Longitude (λ) and geodesic height (h). The ellipsoid parameters are 

expressed in terms of the semi major axis (a) and the flattening (f). These geographic 

coordinates which are then related mathematically to another system of mathematical 

coordinates on a flat/plane surface of a map are known as the rectangular Cartesian grid 

coordinates. 

March 2009 2-1


2.2 MAP PROJECTION MALAYSIA 

2.2.1 Rectified Skew Orthomorphic (RSO) Projection 

Rectified Skew Orthomorphic Projection has been adopted for the Topographic Maps produced by 

the Department of Survey and Mapping Malaysia. The result of this projection is the RSO Grid 

Coordinates. The Datum for this projection is KERTAU (Bukit Kertau Pahang). RSO projection was 

selected to suit the shape of Peninsular Malaysia. The limits of the projection are mainland 

Peninsular Malaysia and the close lying offshore islands. This RSO projection cannot be extended to 

include islands in the South China Sea, nor the East Malaysia states of Sabah and Sarawak. The East 

Malaysia states are covered by a second RSO projection. The Datum for this projection for the land 

below the wind is TIMBALAI (Timbalai Labuan). 

The mathematical theory on which the projection is based is found in the article “The Orthomorphic 

Projection of the Spheroid” by Brigadier M. Hotine CBE, published in the “Empire Survey Review” Vols 

VIII and IX Nos 62-65, particularly para 19 E.S.R. No. 64 of April 1947. 

2.2.2 Cassini Soldner Projection 

Figure 2.2 RSO Grid Projection on Topographic Map 

This projection was used extensively in Great Britain in the 19 th Century where mapping was done by 

the respective counties (Majlis Perbandaran) whose areas are small. However it is not suitable for 

mapping of a nation as the projection is subjected to distortion of scales which increase progressively 

for areas whose distances increase from the central meridian of the ellipsoid. Similarly, the Cassini 

Soldner projection used in Peninsular Malaysia is on a state by state basis (except for the large state 

of Pahang which has 4 zones) by defining a central meridian and origin of projection for each of the 

states. 

Computation of cadastral coordinates for land title survey in Peninsular Malaysia based on the cassini 

soldner projection is very simple. It is based on the concept of selecting a fixed meridian and a point 

2-2 March 2009


on the fixed meridian of the ellipsoid which acts as an origin. The coordinates of any point are then 

found as the length of perpendiculars from the point on the lot of a piece of land to the fixed 

meridian and the distance of the foot of the perpendiculars from the origin point. 

Geographical coordinates controlling cadastral surveys are computed on three separate datums 

namely the ASA datum (Bukit Asa) for the Southern Part of Peninsular Malaysia, the Kertau MRT 

datum for Terengganu, Perak and Kelantan and the Perak Datum (Gunong Hijau Larut) for Perlis, 

Kedah and Penang. However each state adopts its own coordinates system. 

2.2.3 WGS (World Geodetic System) 84 Ellipsoid 

A unified global World Geodetic Reference System for relating the position of any feature or object 

on the surface of the earth become essential in the 1950s for several reasons:- 

• International space science and the beginning of astronautics 

• The lack of inter-continental geodetic information 

• The inability of the large geodetic systems to provide a worldwide geographic coverage 

• Need for universal geographic reference system for global maps used for navigation, aviation 

and geography or surveying 

The new World Geodetic System called WGS 84 is currently the reference system used by the Global 

Positioning System. The WGS 84 originally used the GRS 30 reference ellipsoid but has undergone 

some minor refinements to meet high-precision calculations for the orbits of satellites. However 

these have little practical effect on typical topographic maps. Currently survey works by the 

Department of Survey and Mapping using GPS (Global Position System) is based on WGS 84 

coordinates published by JUPEM (Jabatan Ukur dan Pemetaan) in 1994. 

2.2.4 GDM 2000 or Geocentric Datum Malaysia 2000 

The increasing usage of GPS by surveyors, engineers, navigators and other professionals especially 

those in GIS (Geographic Information System) applications, means that JUPEM has to provide 

geographically referenced map products which are compatible with worldwide usage of GPS without 

having to resort to lengthy computation steps which involves the transformation of coordinates such 

as follows:- 

(Ø λ h) < > (Ø λ h) < > (N, E.) < > (N, E) 

(WGS84) (MRT) (RSO) (Cassini) 

Future cadastral coordinate system will be based on the Geocentric Datum Malaysia 2000 or 

GDM2000. This system will replace the cassini soldner coordinates system mentioned above to 

facilitate the use of GPS. The GPS network which links all the GPS stations to form the Peninsular 

Malaysia Primary Geodetic Network for GDM2000 is depicted below. 

March 2009 2-3


Latitude °N 

Longitude °E 

Figure 2.3 Peninsular Malaysia GPS Network 

2-4 March 2009


REFERENCES 


[2] United States Geological Survey website Map Projection Poster 

egsc.usgs.gov/isb/pubs/MapProjections/projections.html” 

[3] “The Orthomorphic Projection of the spheroid” Brigadier M. Hotine CBE in the Empire Survey 

Review vols VIII and IX Nos 62-65, particularly para 19 E.S.R. no. 64 of April 1947 

March 2009 2-5

CHAPTER 3 TYPES OF SURVEY

Chapter 3 TYPES OF SURVEY 



3 TYPES OF SURVEY ............................................................................................................... 3-1 

3.1 INTRODUCTION .......................................................................................................... 3-1 

3.2 CLASSIFICATION OF SURVEYS ..................................................................................... 3-1 

3.2.1 Geodetic ...................................................................................................... 3-1 

3.2.2 Plane .......................................................................................................... 3-1 

3.2.3 Construction Surveys .................................................................................... 3-1 

3.2.4 Topographic Mapping Surveys ....................................................................... 3-1 

3.2.5 Basic Control (Geodetic) Surveys ................................................................... 3-2 

3.2.6 Satellite Surveys .......................................................................................... 3-2 

3.2.7 Hydrographic Surveys ................................................................................... 3-2 

3.2.8 Land Surveys ............................................................................................... 3-2 

3.2.9 Engineering Surveys ..................................................................................... 3-2 

3.3 SURVEY NETWORKS .................................................................................................... 3-3 

3.3.1 Basic Horizontal Control Network ................................................................... 3-3 

3.3.2 Basic Vertical Control Network ....................................................................... 3-3 

3.4 REAL TIME KINEMATIC (RTK) SURVEY .......................................................................... 3-3 

3.5 LIDAR (Light Detection and Ranging) Airborne Mapping .................................................. 3-4 

3.6 REFERENCES ............................................................................................................... 3-5 

APPENDIX 3A-1 .................................................................................................................... 3A-1 

APPENDIX 3A-2 .................................................................................................................... 3A-2 

March 2009 3-i



3-ii March 2009


3 TYPES OF SURVEY 


Surveying is the science of determining relative positions of points of geographical features on, 

under, or near the earth’s surface. These points may be cultural, hydrographic or terrain features on 

maps, or points needed to locate or layout roads, waterways, air fields or engineering structures of 

all kind 

3.2 CLASSIFICATION OF SURVEYS 

Surveying which can be classed technically or functionally are described below:- 

3.2.1 Geodetic 

A survey in which the figure and size of the mathematically created ellipsoidal shape of the earth is 

considered. It is applicable for large areas and long lines such as topographic mapping on a national 

scale. It is used for the precise location of higher order basic points in a control framework or net for 

controlling other lower order surveys. The Malaysia Primary Geodetic network and the GDM2000 

Datum are described under “Basic Control (Geodetic) Surveys” item 3.2.5 and shown as Fig 2.3 

Peninsular Malaysia GPS Network. 

3.2.2 Plane 

In plane survey the curved surface of the earth is assumed to be flat. Currently cadastral survey for 

Issue Document of Title under the provision of the National Land Code Malaysia (Act 56 of 1965) is 

based on plane coordinates. For small areas, precise results may be obtained with plane-surveying 

methods, but the accuracy and precision of such results will decrease as the area surveyed is 

progressively increased in size. This is reflected in the need for each of the states in Peninsular 

Malaysia to have its own plane coordinate system except the very large state Pahang which has 4 

zones. 

3.2.3 Construction Surveys 

These surveys are conducted to obtain data essential to plan, design and estimate costs to locate or 

provide the layout points for implementing the construction of engineering structures. These surveys 

normally cover relatively small sites where the use of plane surveying techniques is adequate. 

3.2.4 Topographic Mapping Surveys 

Topographic survey involves both air survey and field survey activities. Topographic surveys are 

conducted to establish horizontal and/or vertical positions of points which are then linked to similar 

distinctly identifiable points captured on aerial photograph for use by photogrammetric interpreters 

to compile topographic maps using computer aided mapping systems. Since the control stations are 

usually distributed over comparatively large areas their relative positions are determined by using 

point positioning by satellite techniques. Currently satellites from the GPS (Global Positioning 

System) which are being utilized globally are also widely used in Malaysia. 

March 2009 3-1


3.2.5 Basic Control (Geodetic) Surveys 

Basic control survey provides positions, horizontal and or vertical, of geographic points on a terrain in 

a control framework to which supplementary surveys are adjusted. Most of these basic controls are 

limited to fit national mapping requirements and cannot be applied internationally. In Malaysia, 

these points are contained in two control network based on two local geodetic datum namely the 

Malayan Revised Triangulation (MRT) network for Peninsular Malaysia (West Malaysia) and the 

Borneo Triangulation 1968 (BT68) network for Sabah and Sarawak (East Malaysia). 

However, with the advent of new technologies such as the Global Positioning System (GPS) and 

Unified Geographic Information System (GIS) over large areas, the existing MRT and BT68 network 

have become outdated. A new Geocentric Datum of Malaysia (GDM2000) which fits into the global 

geodetic framework has been introduced to eventually replace the MRT and BT68. The GDM2000 

datum contains the Peninsular Malaysia Primary Geodetic Network (PMPGN) of permanent GPS 

Stations established in 1998 for geodetic and scientific purposes. A similar East Malaysia Primary 

Geodetic Network (EMPGN) is being established. 

3.2.6 Satellite Surveys 

Satellite surveys employ the use of artificial earth satellites as a means of extending geodetic control 

systems. These positioning of points on the ground in a geodetic control system are being conducted 

using artificial earth satellites in the Global Positioning System (GPS) for long line surveys where the 

distance between stations is a few hundred kilometers apart. They are used for conducting 

worldwide surveys for intercontinental, inter-datum and inter-island geodetic ties. Topographic and 

basic control surveys are frequently conducted with satellite surveys. Special project instructions are 

written to detail methods, techniques, equipment and procedures to be used in these surveys. 

3.2.7 Hydrographic Surveys 

A survey made in relation to any considerable body of water, such as a strip of part of the sea along 

the coast, a bay, harbour, lake or river for the purpose of determination of channel depths for 

navigation, location of rocks, sand bars, and in the case of rivers for flood mitigation control, hydroelectric 

power generation, navigation of boats, water supply and water storage. 

3.2.8 Land Surveys 

Land surveying embraces survey operations to locate and monument the boundaries of a property to 

meet the requirement of Land Laws relating to land and land tenure in the National Land Code (Act 

56 of 1965). In the case where alienated land is acquired for construction works such as flood 

mitigation projects land survey has to be conducted to meet the requirement of the Land Acquisition 

Act. Land survey is commonly referred to as Cadastral Survey. 

3.2.9 Engineering Surveys 

It is executed for the purpose of obtaining information which is essential for planning an engineering 

project or proposed development and estimating its cost. The survey information may, in part, be in 

the form of an engineering survey map. 

3-2 March 2009


3.3 SURVEY NETWORKS 

Horizontal and vertical survey control within a country like Malaysia was established by a network of 

control arcs, which are all referenced to a single datum and are therefore linked in position and 

elevation to each other, regardless of their distance apart. These networks for topographic mapping 

are referenced to the KERTAU Datum for the Malayan Revised Triangulation (MRT) network in 

Peninsular Malaysia and the TIMBALAN Datum for the Borneo Triangulation 1968 (BT68) network in 

the Sabah and Sarawak states of East Malaysia. 

3.3.1 Basic Horizontal Control Network 

The horizontal control for mapping was established by connecting a mixed series of stations 

(geodetic, primary, secondary and tertiary stations) by a combination of precise electronic distance 

measuring techniques (Geodimeter) and first order astronomical observation to form the Malaysian 

geodetic net covering Peninsular Malaysia. The stations in the network were then transformed into 

the RSO coordinates system. This network is being replaced by the GDM2000 network, shown in Fig 

2.3, which has been established using GPS satellite point positioning techniques to fit it into a global 

geodetic framework. This network is termed Malaysia Primary Geodetic Network (PMPGN) and the 

East Malaysia Primary Geodetic Network (EMPGN). 

3.3.2 Basic Vertical Control Network 

This control was established to provide orthometric (mean sea level) heights in the national height 

system in the configuration of leveling networks. The datum for orthometric leveling in Peninsular 

Malaysia is Bench mark No. B0169 Height 3.863 metres above Mean Sea Level (MSL) located at the 

back of the tide gauge station on Warf No. 25 North Port, Port Klang. Hydrographic survey for 

design of marine structures may require the heights to be tied to the Chart Datum used in Nautical 

Charts. In such situations the Orthometric (Mean Sea Level) heights relative to the Chart Datum 

available from the Hydrographic Division of the Royal Malaysia Navy has to be obtained. Fig 4.1 

Survey Datum shows the relationship between the Chart Datum and Land Survey Datum. 

3.4 REAL TIME KINEMATIC (RTK) SURVEY 

The Geodesy Section, Department of Survey and Mapping Malaysia provide Real Time Kinematic 

(RTK) Virtual Reference Station (VRS) technique which extends the use of RTK to the whole of 

Peninsular Malaysia by the establishment of a network containing GPS reference stations over the 

whole of Peninsular Malaysia. This service, which attracts a standard fee, is provided by the Malaysia 

Real-Time Kinematic GPS Network System (MyRTnet), for users to conduct dynamic GPS Survey to 

meet applications below:- 

• Geomatics 

• Deformation Monitoring 

• Scientific Research 

• Surveying 

• Construction 

• Navigation 

• Mapping and GIS (Geographic Information System) 

• Location Based Services 

RTK VRS networking exploits the concept of all users sharing a common GPS coordinate control 

framework and it significantly reduces systematic errors and extends the operating range with 

improved accuracy requiring less time. It is surveying where users do not have to set-up their own 

base stations 

March 2009 3-3


Appendix 3A-1 shows in general the concept on functioning of the RTK network together with cellular 

phone (gsm) communication to obtain the geographical position of a map or engineering feature to 

an accuracy of +/- 2 to 3 cm. 

3.5 LIDAR (Light Detection and Ranging) Airborne Mapping 

Light Detection and Ranging (LIDAR) is an airborne mapping technique which uses Laser to measure 

the distance between the aircraft and the terrain of the ground. Airborne LIDAR systems can broadly 

be classified into 3 main types: Wide Area Mapping using fixed wing aircrafts, Corridor Mapping 

Systems mounted on helicopters and bathymetric mapping systems using either one of these two 

airborne platforms. 

A typical airborne LIDAR system coupled with a Global Positioning System (GPS) and an Inertial 

Navigation System (INS) allow the user to capture geo-referenced “Points” of ground features to 

produce highly accurate Digital Elevation Models (DEMs) either day or night in a variety of weather 

conditions. The LIDAR system acquires data along a corridor that can be up to 600 metres wide. 

These very accurate elevation data have a variety of uses, such as the generation of contour lines, 

beach profiles and modeling terrain for 3D applications. 

Data acquired using LIDAR systems are often used in conjunction with data from other remote 

sensing instruments; including spectral and thermal imaging system and high resolution video and 

digital aerial cameras to produce digitally rectified images or orthophotographs. More information on 

LIDAR is contained in item 4.17 and Appendix 3A-2. 

3-4 March 2009


REFERENCES 




[3] GDM2000 Geodesy Section, Department of Survey and Mapping website 

http://geodesi.jupem.gov.my 

March 2009 3-5


APPENDIX 3A-1 

Illustration on Point Positioning for using Satellite and RTK (Real Time Kinematic) Networking 

JUPEM - Jabatan Ukur Dan Pemetaan Malaysia (Department of Survey and Mapping 

Malaysia) 

MyRTKnet - Malaysia Real Time Kinematic GPS network system control center 

RTCM - Radio Technical Commission for Maritime Services (RTCM) Standard for mobile 

phone communication to enable the field surveyor to obtain the real time position 

of a point to an accuracy of +/- 2 to 3 cm from myRTKnet 

JUPEM GPS reference Station 

stations 

- A GPS station within the JUPEM Network of RTK GPS reference 

3A-1 March 2009


APPENDIX 3A-2 

IMU 

Airborne LIDAR System 

LIDAR (Light Detection and Ranging) Airborne System comprising 

• Laser Scanner 

• GPS (Global Positioning Satellite) Receiver 

• IMU (Inertial Measurement Unit) 

March 2009 3A-2



3A-3 March 2009

CHAPTER 4 REFERENCES ON GEOMATICS AND LAND SURVEY 

SERVICES

Chapter 4 REFERENCES ON GEOMATICS AND LAND SURVEY SERVICES 



List of Figures ........................................................................................................................ 4-iii 

4.1 INTRODUCTION .......................................................................................................... 4-1 

4.2 POINT POSITIONING OF A FEATURE USING SATELLITE AND RTK ........................................ 

(JADUAL 2001 ITEM 1.4). ............................................................................................. 4-1 

4.3 PLANIMETRIC (HORIZONTAL TRAVERSING) CONTROL AND CONNECTION .......................... 

(JADUAL 2001 ITEM 1.5). ............................................................................................. 4-1 

4.4 HEIGHT (VERTICAL) CONTROL AND CONNECTION (JADUAL 2001 ITEM 1.6). .................. 4-2 

4.5 LEVELING BENCH MARKS (BM) OR MONUMENTATION (JADUAL 2001 ITEM 1.7) .............. 4-3 

4.6 TOPOGRAPHICAL SURVEY (JADUAL 2001 ITEM 2.10 AND ITEM 7.9) ................................ 4-4 

4.7 GRID SURVEY (JADUAL 2001 ITEM 7.9.2 AND ITEM 2.2 IN KEMENTERIAN ........................... 

KEWANGAN KHAZANAH MALAYSIA LETTER REFERENCE (K&B)(8.09)735/3/1 JD.3(13) ........... 

DATED 13 TH JANUARY 1984). ...................................................................................... 4-4 

4.8 SETTING-OUT SURVEY (JADUAL 2001 ITEM 8.10 AND 8.13). .......................................... 4-4 

4.9 SURVEY OF EXISTING WATERWAYS, CANALS AND DRAINS ................................................ 

(JADUAL 2001 ITEM 8.11 AND 3.10.2) .......................................................................... 4-4 

4.10 STRIP SURVEY TO MAP DETAILS AND SPOT LEVELS (JADUAL 2001 ITEM 4.9 AND 8.9) .... 4-4 

4.11 PREPARATION OF LAND ACQUISITION PLANS (JADUAL 2001 ITEM 8.14 & 1.11 ................... 

AND REGULATION 1991 ITEM 3(B). .............................................................................. 4-5 

4.12 EFFECT OF ADVANCE OR RETREAT OF THE BED OF ANY RIVER OR SEA .......................... 4-5 

4.13 TRANSFORMATION OF COORDINATES AND MAP PROJECTIONS IS NEEDED DUE ................. 

TO THE USE OF VARIOUS GEOGRAPHIC REFERENCE SYSTEMS (JADUAL 2001 ..................... 

ITEM 8.16 AND 1.13). .................................................................................................. 4-5 

4.14 AIR SURVEY MAPPING TECHNIQUE FOR PRODUCING ENGINEERING SURVEY PLANS 

(JADUAL 2001 ITEM 11) .............................................................................................. 4-6 

4.14.1 Limitation of Air Survey ............................................................................... 4-7 

4.15 HYDROGRAPHIC SURVEY FOR TERRITORIAL WATERS AND INLAND WATER BODIES (JADUAL 

2001 ITEM 14 PART V) ................................................................................................. 4-7 

4.16 LOCATING OF CROSS-SECTION PROFILES FOR HYDRAULIC ENGINEERING (JADUAL 2001 

ITEM 14.9 PART V) ...................................................................................................... 4-8 

4.16.1 Mixed Survey Methods ................................................................................ 4-8 

4.16.2 Guidance to Surveyors on Cross-Section Locations......................................... 4-8 

4.16.3 Guidelines on Locating Cross-Sections .......................................................... 4-8 

4.16.4 Additional Guidelines on Cross-Section Profiles .............................................. 4-9 

4.16.5 Cross-Sections Adjacent to Bridges or Culverts (Jadual 2001 Item 3 Part I) ... 4-10 

4.17 LIDAR (LIGHT DETECTION AND RANGING) AIRBORNE MAPPING .................................. 4-10 

4.18 REFERENCES ............................................................................................................. 4-12 

APPENDIX 4A-1 .................................................................................................................... A4-1 

March 2009 4-i


APPENDIX 4A-2 .................................................................................................................. A4-45 

APPENDIX 4A-3 .................................................................................................................. A4-57 

APPENDIX 4A-4 .................................................................................................................. A4-61 

APPENDIX 4A-5 .................................................................................................................. A4-64 

APPENDIX 4A-6 .................................................................................................................. A4-69 

APPENDIX 4A-7 .................................................................................................................. A4-72 

4-ii March 2009




4.1 Survey Datum 4-3 

4.2 Typical Cross-Section Configuration 4-9 

4.3 Cross-Section Locations at a Bridge or Culvert 4-10 

March 2009 

4-iii



4-iv March 2009


4 REFERENCES ON GEOMATICS AND LAND SURVEY SERVICES 


Guides to components of surveying which are important for ascertaining cost estimates and 

specifying scope of survey works in the planning of any Department of Irrigation and Drainage 

project are currently guided by contents in the following references. Updates which are issued from 

time to time should be applied where relevant in the future to these references. 

a. Kelulusan Kadar Baru Pengiraan Kos Perkhidmatan Perunding Bidang Ukur Tanah bagi 

Projek-Projek Kerajaan. Perbendaharaan Kementerian Kewangan Malaysia letter reference 

S(K&B)(8.09)735/9-24 Sj.5.Jld.3 (11) dated 29 th March 2005. 

b. Jadual Fee Ukur Kejuruteraan 2001 (Pindaan Kepada Jadual Fee Ukur Kejuruteraan 1980). 

Please see Appendix 4A-1. 

c. Peraturan-peraturan Jurukur Tanah Berlesen (Pindaan) 1997 (Kadar Bayaran Upah Ukur 

untuk Ukuran Hakmilik) Akta Jurukur Tanah Berlesan 1958 P.U. (A) 169. Please see 

Appendix 4A-2. 

d. Surat Perkeliling Perbendaharaan Bil.8 Tahun 2006 on Peraturan Perolehan Perkhidmatan 

Perunding reference S/K.KEW/PK/1100/000000/10/31 Jld.21 (5) dated 6 th November 2006. 


e. Chapter 17 River Engineering and Channel Stabilization Surveys EM1110-2-1003 US Army 

corps of Engineers Hydrographic Survey Manual. 

f. BQ Example - Cost Estimate for Survey of Existing Route of Waterways, Canals and Drains. 


g. BQ Example - Cost Estimate for Hydrographic Survey of Territorial Waters and Inland Water 

Bodies. Please see Appendix 4A-5. 

4.2 POINT POSITIONING OF A FEATURE USING SATELLITE AND RTK (JADUAL 

2001 ITEM 1.4). 

The Global Positioning System (GPS) is currently the only fully functional Global Navigation Satellite 

System (GNSS). Utilizing a constellation of at least 24 medium Earth Orbit Satellites that transmit 

precise microwave radio signals, the system enables a GPS receiver to determine the Position of a 

point or location on or above the surface of the earth. The GPS radio receiver has become a widely 

used aid to navigation worldwide and a useful tool, among many others, map making and Land 

Surveying. GPS equipment used by surveyors incorporates techniques and augmentation methods to 

improve accuracy and error sources inherent to operation of GPS. Example of augmentation systems 

includes Differential GPS or RTK (Real-Time Kinematic) surveying illustrated at Appendix 3A-1. 

In Malaysia RTK survey service for a fee is provided by logging on to myRTKnet located at the 

Geodesy Section of the Department of Survey and Mapping. 

4.3 PLANIMETRIC (HORIZONTAL TRAVERSING) CONTROL AND CONNECTION 

(JADUAL 2001 ITEM 1.5). 

Planimetric cntrol and connection is a technique used for determining the relative horizontal positions 

(x, y coordinates) of cultural, hydrographic or terrain features for mapping or points needed to plan 

and subsequently locate positions or layout accurately bunds, canals, soil investigation boreholes, 

roads, waterways and drainage structures, of all kinds. It comprises a series of points on features 

surveyed. Hence it comprises: 

March 2009 4-1


a. Connection to Survey Department Horizontal Datum which provides scale, position and 

azimuth control for establishing boundary marks shown on land title survey or cadastral 

survey plans to meet issue document of titles to land based on the Cassini Soldner 

projection. 

b. And then the proposed or existing routes or alignment, identified by the Department of 

Drainage and Irrigation. Issue to be considered here are land with various category of titles 

and ownership which have to be obtained from the Cadastral Division Department of Survey 

and Mapping, the Land Office and sometimes direct objection from the affected land owner 

himself. 

4.4 HEIGHT (VERTICAL) CONTROL AND CONNECTION (JADUAL 2001 ITEM 1.6). 

Height controls and connection to determine the spot level of a feature includes: 

a. Connection to Survey and Mapping Department Bench Marks (BM) based on the Land Survey 

Datum (LSD) and now known as the National Vertical Geodetic Datum (NGVD) which is 

located at a tide gauge station sited in Port Klang, Selangor 

b. Connection to the CHART DATUM which is traditionally referred to as the Admirably Chart 

Datum. These datums are located at Tidal Stations, mainly jetties or ports along the coast. 

Appendix 3A-1 attached contains a list of the existing Tidal Stations. 

c. Occasionally connection to both the LSD and the Admirably Chart Datum has to be related 

for marine navigation structures such as a fishing jetty or port. An example of this is 

depicted in the diagram below which shows the Chart Datum is 1.7m below the Land Survey 

Datum. 

d. And then along the proposed or existing routes or alignment identified by the Drainage and 

Irrigation Department 

4-2 March 2009


TIDAL REFERENCE FOR PULAU SIBU 

31.847 m (BM S1150) 

Above L.S.D 

00m (L.S.D)/(M.S.L) 

32.547m (BM S1150) 

Above Chart Datum 

1.700m 

NAUTICAL OR ADMIRALTY CHART DATUM 

-1.700m (Chart Datum) 

00 (Chart Datum) 

Chart Datum is 1.700M below Land Survey Datum (L.S.D) at 

Survey Department Bench Mark (BM S1150) 

Figure 4.1 Survey Datum 

4.5 LEVELING BENCH MARKS (BM) OR MONUMENTATION (JADUAL 2001 ITEM 

1.7) 

A Bench Mark is a relatively permanent object natural or artificial, bearing a marked point normally a 

brass bolt set in concrete with a bench mark number inscribed. The elevation or the height of the 

point above or below the Land Survey Datum (LSD) or National Geodetic Vertical Datum (NGVD) has 

to be purchased from Geodesy Section, Department of Survey and Mapping. Establishment of 

subsidiary marks or monuments related to the Department of Survey and Mapping Bench Marks by 

conducting Height Control and Connection Surveys are known as:- 

a. Temporary Bench Marks (TBM) 

• Plan of a TBM marker on Normal surface is shown in Appendix 4A-6. 

• Plan of a TBM marker on hard surface is shown in Appendix 4A-7. 

b. Intersection Point Marks (IP) 

c. Reference Marks (RM) 

March 2009 4-3


4.6 TOPOGRAPHICAL SURVEY (JADUAL 2001 ITEM 2.10 AND ITEM 7.9) 

Topographic Surveys often known as Engineering Surveys are conducted to establish horizontal (x, y) 

and/or vertical (h or z) positions of points of all natural and manmade features to produce a 

geographical details and contour map over a large area. Topographic maps supply a general image 

of the earth’s surface namely roads, rivers, buildings, often the nature of the vegetation, the contour 

together with spot levels and names of various surveyed objects. The main supplier of topographic 

map is the Department of Survey and Mapping Malaysia 

4.7 GRID SURVEY (JADUAL 2001 ITEM 7.9.2 AND ITEM 2.2 IN KEMENTERIAN 

KEWANGAN KHAZANAH MALAYSIA LETTER REFERENCE (K&B)(8.09)735/3/1 

JD.3(13) DATED 13 TH JANUARY 1984). 

This survey is special to projects where the difference in spot levels is very important and critical. It 

is specified for survey of aircraft runway construction or other flat surface. This type of survey is not 

suitable for undulating or hilly area covered by overgrown vegetation. 

4.8 SETTING-OUT SURVEY (JADUAL 2001 ITEM 8.10 AND 8.13). 

This survey, also known as construction setting out survey, is executed before construction works 

can start. The setting comprise x and y coordinates of the following:- 

a. Centre line of proposed route from IP to IP (Intersection Points) 

b. Right of Way (ROW) of the waterway, canal or drain reserve based on the approved precomputation 

plan. 

c. Intersection Points (IP) (Jadual 2001 Item 8.10) 

d. Pegging of positions of Piling Points based on pre-computation plan from engineering layout 

plan 

4.9 SURVEY OF EXISTING WATERWAYS, CANALS AND DRAINS (JADUAL 2001 

ITEM 8.11 AND 3.10.2) 

This survey covers the area within the banks or the designated or gazette reserve for the irrigation 

canal or waterway to show the alignment, longitudinal section and the cross-sections. It also 

includes the area within the specified Right of Way (ROW) shown on the approved pre-computation 

plan. When the reserve is not specified the outer limits of the alignment is within 50m from the 

banks of river or drain or canal. If the water depth of the waterways, drains and canal at the time of 

survey is more than 1 meter then Jadual 2001 item 8.11 specification (viii) and item 3.10.2 applies 

or alternatively Hydrographic Survey for Inland Water Bodies under Paragraphs 4.15 (Jadual 2001 

Item 14 Part V) and 4.16 (Jadual 2001 Item 14.9.1 Part V) is applicable. If the width of the crosssections 

or the intervals is more or less than 50 metres then the fees shall be increased or decreased 

proportionately (specification (vii) Jadual 2001 item 8.11) 

4.10 STRIP SURVEY TO MAP DETAILS AND SPOT LEVELS (JADUAL 2001 ITEM 4.9 

AND 8.9) 

The strip comprises topographic details and spot levels survey of long narrow stretches of areas or 

corridors which are beyond the banks or overbanks and flood plains of a waterway or along the 

coast. 

4-4 March 2009


4.11 PREPARATION OF LAND ACQUISITION PLANS (JADUAL 2001 ITEM 8.14 & 

1.11 AND REGULATION 1991 ITEM 3(B). 

These are approved pre-computation plans based on the Right of Way (ROW) plans or any other 

area to be acquired for the implementation of the project for the proposed waterway, canal or drain. 

On being surveyed the ROW will eventually become the designated drainage reserve alignment to be 

maintained by the Drainage and Irrigation Department. Land acquisition is a socially sensitive, very 

costly, tedious and long drawn process which entails the following:-. 

a. Preparation of Land Acquisition Plans which comprise:- 

• Purchasing certified plans (Pelan Akui) and Cadastral (Standard) Sheets from the 

Department of Survey and Mapping for the compilation/preparation of Land Acquisition 

(LA) Plans. 

• Search for Qualified Titles (Hakmilik Sementara) and Approved LA Plans at the Land 

Office or other Government Department. 

b. Land Acquisition Plans normally compiled on the same scale as the Survey Department 

cadastral sheet shall show:- 

• Lot boundaries with bearings and distances within the surveyed corridor or strip 

(proposed alignment/ROW) 

• Lot numbers of lots to be acquired 

• Lot areas with details on portion to be acquired and the left over balance 

• Status and category of land use and crops 

• Houses and other as-built features affected by the Acquisition 

c. Finalized Land Acquisition Plans are updated from:- 

• Revision/amendment of ROW by consulting engineer 

• Comments by the Department of Drainage and Irrigation 

• Up-to date information on change in status of land received from the Land Office 

• Objection from Land owner during field survey work to demarcate the ROW/alignment of 

the future waterway reserve or alignment. 

d. R.S. (Requisition for Survey) Plan. The approved Pre-computation plan for Land Acquisition 

which is attached to the Requisition for Survey (Permintaan Ukur) letter by the Land Office to 

the Department of Survey Mapping is known as the R. S. Plan. 

4.12 EFFECT OF ADVANCE OR RETREAT OF THE BED OF ANY RIVER OR SEA 

Frequently while conducting survey for Land Acquisition we come across a situation where part a of 

privately owned land along river banks are lost through erosion by the action of flood water. Similarly 

land along the opposite bank, especially on bends, may also gain land through accretion by the 

action of flooding. Such lands, as per provision of Section 49 of the National Land Code (Act 56 of 

1965), shall become State land. 

4.13 TRANSFORMATION OF COORDINATES AND MAP PROJECTIONS IS NEEDED 

DUE TO THE USE OF VARIOUS GEOGRAPHIC REFERENCE SYSTEMS (JADUAL 

2001 ITEM 8.16 AND 1.13). 

Coordinates in a common geographically referenced system is needed to provide information on the 

location of a position of a feature for navigation, point of Interest or geographic information system. 

An example on the request for transformation of coordinates is the experience with the Sungai Muda 

Flood Mitigation Project stretching from Jambatan Merdeka to Kuala Muda where different 

coordinates are being used. 

March 2009 4-5


a. The various coordinates used are:- 

• The plane cadastral coordinates (x and y coordinate) on the Kedah side of Sungai Muda 

is based on the Cassini Solder projection from the Central Meridian of the Everest 

(Modified) Ellipsoid at Gunung Perak 

• The plane cadastral coordinates in Pulau Pinang side of Sungai Muda is based on the 

Cassini Solder projection from the Central Meridian of the Everest (Modified) Ellipsoid at 

Gun Hill. 

• The coordinates required by the Civil and Structural Consultants is that based on the 

RSO (Rectified Skew Orthomorphic) projection. The RSO projection is used for 

Topographic Maps produced by the Department of Survey and Mapping Peninsular 

Malaysia. The RSO projection (Fig. 2.2.2) was selected because of the shape of the area 

to be mapped and the scale distortion which can be tolerated. 

• WGS (World Geodetic System) 84 coordinates (Jadual 2001 item 1.4 Part I). The world 

Geodetic System 1984 coordinates are used when Point Positioning is determined using 

GPS (Global Positioning Satellites). The World Geodetic System (WGS84) the latest 

revision is WGS84 dating from 1984 (last revised in 2004) will be valid to about 2010. A 

unified World Geodetic System based on the WGS84 ellipsoid is essential for several 

reasons:- 

- International space science and astronautics 

- Inter-continental geodetic information 

- Inability of large geodetic systems such as the Rectified Skew Orthomorphic (RSO) 

for Peninsular Malaysia which cannot be extended to include islands in the South 

China Sea nor the East Malaysia State of Sabah and Sarawak; European Datum 

(ED50) and North American Datum (NAD) to provide worldwide coverage to meet 

the need for global or regional maps for navigation, aviation and geography 

b. Eventually when the GDM2000 coordinates system is fully implemented the requirement for 

coordinates transformation may be greatly reduced. GDM2000 is described at item 2.2.4. 

c. The Consulting Civil and Structural Engineers requirement for engineering survey plans to be 

in RSO Coordinates against plans in the respective Land Office in Kedah and Pulau Pinang in 

their respective Cadastral Cassini Solder Coordinates require the mathematical process of 

transformation of coordinates e.g. WGS84 to RSO or Kedah Cassini to RSO or vice versa 

4.14 AIR SURVEY MAPPING TECHNIQUE FOR PRODUCING ENGINEERING SURVEY 

PLANS (JADUAL 2001 ITEM 11) 

The provision here is for the out-put of photo-mosaics and photo-maps over a wide area or long 

corridor using aerial photographs supplied by the Department of Survey and Mapping. If the Survey 

Department aerial photographs are out of date “Jadual 2001 item 12” caters for acquisition of new 

ones by Air Survey methods. The benefits of adopting this approach are:- 

a. Access and Coverage - Aerial images can be obtained of areas that are inaccessible or 

dangerous for ground surveyors due either to unfriendly inhabitant, difficult terrain or a need 

to maintain confidentiality. An accurate survey can then be compiled in comfortable 

surroundings. The approximate width of the corridor covered is 1000m (1km) whereas the 

actual Right of Way (ROW) may be 100m. It provides advance survey information over a 

wider area which can then be narrowed down to the proposed corridor requiring follow-up of 

more detailed field survey works. 

b. Speed and Cost - Due to the high speed of aerial surveys the cost of works is reduced, and 

the final product is available earlier. In addition, the expenses of working away from base 

are reduced, as only the flying crew and some camera operators need travel to the survey 

area. The photomap together with the photo-mosaic will provide a more focused approach to 

the planning and scheduling of the actual field survey works. 

4-6 March 2009


c. Control - The organization and control of the survey is simplified as the bulk of the surveyors 

are working in good stable conditions at base, where they can be easily administered and 

supervised. Those conditions produce high output and quality of work. 

d. Supply - Supply is also simplified as in most cases the aircraft can operate from a commercial 

airport. Any necessary special equipment such as GPS enabled digital camera can be carried 

in the survey aircraft to the area of operations. 

e. Weather - Although low cloud and extensive cloud-cover will prevent photography, only a 

short time is needed to obtain suitable images. The weather is therefore seldom a major 

problem, and once the photographic data has been obtained the survey is unaffected by 

weather conditions. 

4.14.1 Limitation of Air Survey 

Survey Control - In order to relate an air-survey to the area in which the images were taken, it is 

necessary to have precise ground coordinates, both plan and height, of points that can be clearly 

seen on the images and on the ground. Coordinates and a clear description of each point are 

provided by the ground surveyors as control for the aerial survey. Whilst aerial triangulation using 

electronic computers provides a means of distributing additional controls on photographs a certain 

amount of ground control is necessary, and must be provided before the air survey mapping works 

can be commenced. Invert levels below the water surface cannot be ascertained. 

Check - A field check of an air survey is necessary to eliminate errors due to misinterpretation of 

detail. If the survey is at a large scale, completion of hidden detail (under trees, in shadow, etc) 

may be needed. In all cases, names and description must be obtained from ground survey works. 

Administrative work include:- 

a. Arrangement for tasking of aircraft 

b. Application for security clearance and the obtaining of the permit to fly aerial photographic 

mission 

c. Mobilization of personnel and equipments 

4.15 HYDROGRAPHIC SURVEY FOR TERRITORIAL WATERS AND INLAND WATER 

BODIES (JADUAL 2001 ITEM 14 PART V) 

Hydrographic survey provides information and data to support:- 

a. The management of coastal zones 

b. The hydrographic survey of deltaic regions and river months up to two kilometers upstream 

of river mouth 

c. The development of coastal engineering, property, infrastructure projects and activities 

d. The management and development of jetties, ports, harbors and associated maritime 

facilities 

e. The management and development along inland waterways and inland water body 

March 2009 4-7


4.16 LOCATING OF CROSS-SECTION PROFILES FOR HYDRAULIC ENGINEERING 

(JADUAL 2001 ITEM 14.9 PART V) 

4.16.1 Mixed Survey Methods 

Obtaining cross-section profile of stream, adjoining bank and flood plain requires a combination of 

survey methods. Hydrographic sounding surveys performed in the river must be combined with 

conventional topographic, and or photogrammetric surveys in the adjacent over banks and flood 

prone plain. Surveys of the flood plains are usually more efficiently conducted using air survey 

(Digital photogrammetric) methods to create a Digital Elevation Model (DEM). Recently, airborne 

LIDAR (Item 4.17) techniques have been developed to provide DEM of the flood plain. Conventional 

topographic survey methods (leveling and digital/optical total station) will be required to fill in hidden 

areas under cover of vegetation and to ascertain break lines in the final terrain models. 

4.16.2 Guidance to Surveyors on Cross-Section Locations 

Detailed guidance for determining the location and spacing of stream cross-sections is based on the 

recommendations in the US Army Corps of Engineers, Engineers Manual “EM1110-2-1002” and 

EM1110-2-1416”. Surveyors providing input for these studies should be aware of the hydraulic 

considerations that dictate the intended placement and alignment of stream sections. Thus, 

knowledge of the engineering rationale for locating cross-sections profiles is required by field 

surveyors in order to make reasonable adjustments or recommend modification to the project 

engineer to optimize the obtaining of basic field information on the river profile, the adjoining river 

banks and the flood plain. 

4.16.3 Guidelines on Locating Cross-Sections 

Generally (not exhaustive) the locations of Cross-sections for hydraulic modeling should be 

considered are:- 

a. Points where roughness changes abruptly to provide channel roughness information 

b. Closer together in stretches where water surface expands and in bends 

c. Closer together in stretches where the flow of water changes greatly as a result of changes 

in width, depth or roughness 

d. Closer together at wide bends where the lateral distribution of water flow changes radically 

with distance 

e. Closer together in streams of very low gradient at lowlands which are significantly non 

uniform, because the computations are very sensitive to the effects of local disturbances 

and/or irregularities 

f. At tributaries that contribute significantly to the main stem flow. Cross-sections should be 

located immediately upstream and downstream from the confluence on the main river and 

immediately upstream on the tributary 

g. At regular intervals along waterway of uniform cross-section 

h. Above, below, and within bridges at bridge sites including the soffit levels 

i. On large rivers that have average slopes of 0.4 meter to 1.5 meter per kilometer or less, 

cross-section within fairly uniform reaches may be taken at intervals of 1.5 km or more 

j. More closely spaced cross-sections are usually needed to define energy losses in urban 

areas, where steep slopes are encountered, and on relatively narrower streams. On small 

streams with steep slopes it is desirable to take cross-sections at intervals of 500m or less. 

4-8 March 2009


k. Recommended maximum reach lengths (distances between cross-sections) are: (1) 800m for 

wide flood plains and slope less than 0.4m per km, (2) 550m for slopes less than 0.6m per 

kilometer, and (3) 365m for slopes greater than 0.6m per kilometer. In addition, no reach 

between cross-sections should be longer than 75 – 100 times the mean depth for the largest 

discharge, or about twice the width of the reach. The fall of a reach should be equal to or 

greater than the largest of 0.15m or the velocity head, unless the bed slope is so flat that 

the above criterion holds. The reach length should be equal to, or less than, the 

downstream depth for the smallest discharge divided by the bed slope 

Figure 4.2 Typical Cross-Section Configuration 

4.16.4 Additional Guidelines on Cross-Section Profiles 

Field surveyors should also take into consideration the following application when acquiring crosssectional 

data. 

a. Cross-sections are run perpendicular to the direction of flow at intervals along the river. The 

“reach length” is the distance between cross-sections. Flow lines are used to determine the 

cross-section orientation. The hydraulic engineer will provide these orientations to the 

surveyor. 

b. The cross-section should be referenced to the stream thalweg (deepest part of the channel) 

and by river kilometers measured along the thalweg. From this the reach lengths (distance 

between cross-sections) is computed. End points on the cross-section should be 

geographically coordinated using the local State Plane Cassini Soldner Coordinate System. 

c. End station elevations. The maximum elevation of each end of a cross-section should be 

higher than the anticipated maximum water surface level. 

March 2009 4-9


d. Local irregularities in bed surface. Local irregularities in the ground surface such as 

depressions or rises that are not typical of the reach should not be included in the crosssectional 

data. 

e. Bent cross-sections. A cross-section should be laid out on a straight line if possible. 

However, a cross section should be bent if necessary to keep it perpendicular to the 

expected flow lines. 

f. Avoid intersection of cross-sections. Cross-sections must not cross each other. Care must 

be taken at river bends and tributary junctions to avoid overlap of sections. 

g. Inclusion of channel control structures. Channel control structures such as bunds or wing 

dams should be shown on the cross-section, and allowances in cross-sectional areas and 

wetted perimeters should be made for these structures. 

4.16.5 Cross-Sections Adjacent to Bridges or Culverts (Jadual 2001 Item 3 Part I) 

Cross-sections need to be denser near bridges and culverts in order to analyze the flow restriction 

caused by these structures. A guide on the locations of cross-sections is shown below. 

RIVER 

CONTRACTION 

W 

CH 001 

UP STREAM 

CH 002 

CROSS 

SECTION 

W 

L 

BRIDGE/CULVERT 

CH 003 

EXPANSION 

4XL 

DOWN 

STREAM 

L – Length of abutment 

W- Span of bridge 

CH 004 

Figure 4.3 Cross-Section Locations at a Bridge or Culvert 

4.17 LIDAR - LIGHT DETECTION AND RANGING AIRBORNE MAPPING 

Information on this aspect of surveying, which was described and illustrated in Appendix 3A-2. and in 

item 3.5 earlier can be found from the web by keying in the following:- 

a. LIDAR technologies 

b. us army corps of engineers hydrographic survey manual (Click item EM1110-2-1003) 

LIDAR technology which is similar to radar is an airborne laser mapping technique. A typical airborne 

LIDAR system is coupled with a Global Positioning System (GPS) to determine aircraft position and 

an Inertial Navigation System (INS) or Inertial Measuring Unit (IMU) to determine the constantly 

changing aircraft attitude. Appendix 3A-2 shows the operation of a typical LIDAR using a fixed wing 

4-10 March 2009


aircraft. 

With LIDAR, highly accurate digital elevation (DEM) and digital terrain (DTM) models or elevation 

contours can be generated immediately well before any aerial photography is processed, ground 

control is acquired and photogrammetric mapping is performed. LIDAR can capture data with 

accuracies of 5 to 20 centimeters to meet modeling efforts day and night in a variety of weather 

conditions. 

However integrating LIDAR data with photogrammetric data from air survey often yields better endresults 

since shorelines frequently have heavy ground vegetation cover and mapping goals are 

frequently 1 to 2 feet (30cm to 60cm) contours. In other words, combined LIDAR – Air 

Survey/Photogrammetric Mapping provides a more realistic depiction of the terrain and ensures 

desired map accuracies will be maintained by providing an independent check. 

Airborne LIDAR system can be broadly classified into 3 main types: wide area mapping systems 

flown from fixed wing aircraft, Corridor mapping systems from helicopters and Bathymetric mapping 

systems flown from either one of the platform. 

March 2009 4-11


4.18 REFERENCES 

[1] US Army Corps of Engineers website is accessible by keying in “us army corps of engineers 

hydrographic survey manual” then click “EM 1110-2-1003 Title: Engineering and Design – 

Hydrographic Survey” 





4-12 March 2009


APPENDIX 4A-1 

SCHEDULE ‘C’ – TREASURY APPROVED RATE 

(JADUAL FEE UKUR KEJURUTERAAN 2001) 

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APPENDIX 4A-2 

SCHEDULE ‘D’ – AKTA JURUKUR TANAH BERLESEN 1958 

P.U. (A) 169. 

(Relevant Pages Only) 

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4A-56 March 2009


APPENDIX 4A-3 

MINISTRY OF FINANCE LETTER 

ON 

MACRES (MALAYSIAN CENTRE FOR REMOTE SENSING) 

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APPENDIX 4A-4 

BQ EXAMPLE – COST ESTIMATE FOR 

SURVEY OF EXISTING ROUTE OF WATERWAYS 

CANALS AND DRAINS 

(2.4.9) 

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4A-62 March 2009


APPENDIX 4A-4 

Item Description Estd Actual Unit Rate Agreed Actual Remarks/ 

Qty Qty (RM) Amount Amt Try Item 

(a) (b) (c) (d) (e) (f) (g) (h) (I) 

1 Waterways and hydrographic survey 

of Muda River (Tidal River) 

1.1 Preparatory work 1 P/day 743.00 743.00 8.1 

1.2 Mobilization & demobilization 18 P/day 743.00 13,374.00 8.2 

(Six Field Parties) respectively 

1.3 Planimetric control for as-built 33.8 km 2,500.00 84,500.00 8.3 

both banks (length 33.8 km) 

Kedah & Penang 

1.4 Height control from existing 13 km 743.00 9,659.00 8.4 

bench mark (misclosure check) 

1.5 Strip survey with details of existing 

tidal waterway (2 x 250m over banks 

+ 100m waterway )waterway 

a) Alignment survey (4.25 x RM243) 13 km 3,157.75 41,050.75 8.11 

b) Cross-section survey at 100m interval 53.4 km 5,944.00 317,409.60 8.11/3.10.2 

c) Long-section survey at 13 km 5,944.00 77,272.00 8.11/3.10.2 

100m interval 

1.6 Establishment of TBMs (Monumentation) 11 No 148.50 1,633.51 8.5 

1.7 Site Survey & preparatory works 

(minimum fee) 

BQ EXAMPLE - COST ESTIMATE FOR SURVEY OF EXISTING 

ROUTE OF WATERWAYS CANALS AND DRAINS 

a) Site No. 1 min 1 1,486.00 1,486.00 } 

b) Site No. 2 min 1 1,486.00 1,486.00 } 

c) Site No. 3 min 1 1,486.00 1,486.00 } 7.10 & 8.12 

d) Site No. 4 min 1 1,486.00 1,486.00 } 

e) Barrage min 1 1,486.00 1,486.00 } 

f) Jambatan Merdeka min 1 1,486.00 1,486.00 } 

1.8 Others 

1.9 Re-imbursable cost for purchase 8.8.8.1.10 

of Revenue Sheet (Std Sheets) 

CPs, hire of boat and travelling 

expenses 

1.10 Land Acquisition Plans 

a) Preparatory work 1 P/day 743.00 743.00 3.11/1.11.1 

d) Search at Land Office 16 hour 10.00 160.00 3.11/1.11.3 

c) Computation Plan 704 lot 20.00 14,080.00 3.11/1.11.4 

569,540.86 

2 Add 5% Government Service Tax 113,908.17 

683,449.03 

3 Supply of Land Acquisition Plans 

Penang 40 sets @ 10 plan/set 40 100 plan 10.00 400.00 1,000.00 8.17/1.14.3 

Kedah 40 sets @ 10 plan/set 40 100 plan 10.00 400.00 1,000.00 

Note: 

a) Alignment survey comprise location of form lines of the waterway 

Estimated Total 

b) Rate of 8 party day per kilometre if the depth of water is more that 1 metre 

(Specification (viii) Item 8.11 Jadual 2001) for cross-section and Longitudinal section 

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APPENDIX 4A-5 

BQ EXAMPLE – COST ESTIMATE FOR HYDROGRAPHIC SURVEY 

OF TERRITORIAL WATERS AND INLAND WATER BODIES 

(2.4.16) 

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APPENDIX 4A-5 

BQ EXAMPLE - COST ESTIMATE FOR 

COASTAL AND WATERWAYS HYDROGRAPHIC SURVEY 

Item Description Estd Actual Unit Rate Agreed Actual Remarks/ 

Qty Qty (RM) Amount Amt Try Item 

(a) (b) (c) (d) (e) (f) (g) (h) (I) 

1 Mobilization & demobilization of 3 P/day 743.00 2,229.00 14.1 

topographic survey equipment 

2 Planimetric control and connection in 10 km 2,500.00 25,000.00 14.2 

built up area 

3 Height control and connection 10 km 743.00 7,430.00 14.3 

4 Topographic strip survey with details 100 ha 99.00 9,900.00 14.8 

100m x 10 km coastal strip 

5 Bathymetric (Off Shore) Profiling 

a) Profiles at 50m 90 km 297.20 26,748.00 14.9.2 

b) Profiles at more than 100m 30 km 371.50 11,145.00 14.9.2 

c) Extended hydrographic survey 9 km 222.90 2,006.10 14.9.2 

up-stream at 25m intervals 

6 Direct Reading of Tide Pole 11 No 148.50 1,633.51 14.10.2 

a) Installation of Tide Pole 1 no 900.00 900.00 

b) Tidal observation 2 P/day 743.00 1,486.00 

88,477.61 

7 Add 5% Government Service Tax 17,695.52 

106,173.13 

8 Boat 

a) Mobilization 1 no 600.00 600.00 

b) Rental 5 P/day 300.00 1,500.00 

Estimated Total 108,273.13 

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APPENDIX 4A-6 

TEMPORARY BENCH MARK (TBM) 

MARKERS ON NORMAL SURFACE 

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5mm Ø Drilled Center 

End Cap Sealed to Pipe 

150 

Concrete 

600 

300 Projection 

300 

50mm Ø G.I Pipe 

IP. 28 

JPS 

NAME OF 

SURVEYOR 

Figures Engraved on 

Concrete 

TBM MARKER ON NORMAL SURFACE 

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APPENDIX 4A-7 

TEMPORARY BENCH MARK (TBM) 

MARKERS ON HARD SURFACE 

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30 50 50 

Cement/Mortar Mix 

10mm Ø Rivet 

6mm Thk. Galvanised 

Steel Plate 

50 

4 Nos. 150mm Galvanised 

Steel Nails driven into 

concrete or hard surface 

(except pavement) 

300 

TBM 19 

300 

JPS 

NAME OF 

SURVEYOR 

30 

10mm Ø Rivet 

Engraved 

Figures 

30 

TBM MARKER ON HARD SURFACE 

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CHAPTER 5 GEOGRAPHIC INFORMATION SYSTEM (GIS)

CHAPTER 5 GEOGRAPHIC INFORMATION SYSTEM 

(GIS)

Chapter 5 GEOGRAPHIC INFORMATION SYSTEM (GIS) 



5.1 INTRODUCTION .......................................................................................................... 5-1 

5.2 MORE ON GIS INFORMATION ....................................................................................... 5-1 

5.3 REFERENCES ............................................................................................................... 5-2 

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5-ii March 2009



5 GEOGRAPHIC INFORMATION SYSTEM (GIS) 

There is an old common saying that “a picture is worth a thousand words”. Today with GIS we can 

choose to have not only the picture which is very often the digital topographic map but also the 

thousand words which is interlinked with the geographically referenced features depicted on a map 

through feature codes. A GIS is a computerized system capable of capturing, storing, analyzing and 

displaying geographically referenced information; that is data of map features identified according to 

location. Traditionally such a graphic picture is depicted on cartographically enhanced topographic 

maps (USGS website on geographic information system http//:egsc.usgs.gov/isb/pubs/gis_poster/). 

GIS tools and methods can be used for environmental studies, water resource management for 

agriculture, flood mitigation development planning or scientific investigation. A GIS may allow flood 

emergency planners to easily calculate flood emergency response times during a flood season. 

Together with cartography a component of topographic mapping, remote sensing, global positioning 

systems, photogrammetry, and geography; GIS has evolved into a discipline with its own research 

base known as Gographic Information science 

An example on the usefulness of GIS technology development is the possibility of combining 

agricultural or land records, hydrography; which include rainfall data, to determine which river will 

carry certain levels of soil erosion sediment runoff. 

Having gone through the above it is hoped the user of this manual can now make use of the link 

provided by the Malaysian Centre for Geospatial Data Infrastructure [2] (MaCGDI) Ministry of Natural 

Resources and Environment (NRE) website http://www.mygeoportal.gov.my to contact various other 

departments to share experience and ideas on creating geospatial information. 

5.2 MORE ON GIS INFORMATION 

More information which is listed below can be obtained from the USGS website mentioned in item 

5.3 References. 

• How does a GIS work? 

• Data Capture 

• Data integration 

• Map projection and registration 

• Data structures 

• Data modeling 

• What’s special about a GIS? 

• Framework for cooperation etc. 

March 2009 5-1


5.3 REFERENCES 



5-2 March 2009

CHAPTER 6 CHECKLIST FOR TERRAIN FEATURES

CHAPTER 6 CHECKLIST FOR TERRAIN FEATURES

Chapter 6 CHECKLIST FOR TERRAIN FEATURES 



6.1 SURVEY SERVICES ....................................................................................................... 6-1 

6.2 LAND ACQUISITION BASE PLAN. .................................................................................. 6-1 

6.3 GROUND MARKERS ...................................................................................................... 6-1 

6.4 INDUSTRY .................................................................................................................. 6-1 

6.5 ROAD FURNITURE, SERVICES AND UTILITIES ............................................................... 6-2 

6.6 BOUNDARY FEATURES ................................................................................................. 6-2 

6.7 BRIDGE SITE ............................................................................................................... 6-2 

6.8 RAILWAYS .................................................................................................................. 6-2 

6.9 SURVEY CONTROL ....................................................................................................... 6-3 

6.10 PLANTATIONS, TREES AND RECREATIONAL AREAS ........................................................ 6-3 

6.11 SLOPES AND EARTHWORKS ......................................................................................... 6-3 

6.12 WATER AND DRAINAGE ............................................................................................... 6-3 

6.13 REFERENCES ............................................................................................................... 6-4 

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6-ii March 2009


6.1 SURVEY SERVICES 

6 CHECKLIST FOR TERRAIN FEATURES 

The survey services to be provided by the surveyor are as listed below and as detailed: - 

a. Discussion with relevant authorities such as JKR, JPS, Survey Department, Local Authority 

and Land Office before the physical commencement of work on site 

b. Consultation with the Superintending Officer (SO) or SO’s Representative and obtain 

Instructions 

c. Study all relevant information, maps and plans provided and obtaining all necessary 

additional topographic maps, certified plans, revenue sheets, data and other information for 

the proper execution of the works 

d. Preparation of topographic survey plans 

e. Field survey to pick up details according to format required 

f. Compiling, processing and preparing data and CAD plot of survey plan in accordance to 

format required 

g. In carrying the work, the surveyor shall attempt to obtain permission prior to entry into 

private land, cemeteries and property of other relevant authorities 

6.2 LAND ACQUISITION BASE PLAN. 

The drawing shall show the following: 

a. Name of districts and mukims 

b. Lot boundaries and lot numbers 

c. Existing total lot areas computed based on coordinates 

d. Land use indicating type of cultivation etc. 

e. Type of building indicating permanent or semi permanent and usage 

f. The existence of burial ground if any within the survey corridor 

g. All other relevant details as instructed by client or as desired by the government 

h. Land lots that are partially within the mapping area shall, where possible, be presented 

showing the whole area of the lot 

6.3 GROUND MARKERS 

The surveyor shall supply two copies of the following results to the client on completion of field work 

and adjustment: 

a. Schedule of all Permanent Ground Markers (TBM’s and RM’s) giving the reference numbers, 

coordinates and heights 

b. Descriptions of Permanent Ground Markers giving the types of marker constructed and 

location 

c. Diagrams of the horizontal control net showing the connection between Permanent Ground 

Markers 

d. Diagrams of the leveling (height control) net indicating the connection between Permanent 

Bench Marks 

6.4 INDUSTRY 

a. Tanks 

b. Valve chambers 

c. Transformers (boundary fences and building lines) 

d. Electricity sub-station, boxes and switch boxes (boundary fences and building lines) 

e. Pylon lines (indicate levels at lowest point at sag and at pylon towers) 

f. Pylon bases 

g. Pylon reference numbers and 

March 2009 6-1


h. Telegraph lines 

6.5 ROAD FURNITURE, SERVICES AND UTILITIES 

a. Km post (value to be noted) 

b. Guardrails 

c. Bus stops 

d. Lamp posts 

e. Telecom poles 

f. Electricity poles 

g. Road signs 

h. Large road signs (with minimum 2 posts only) 

i. Hoardings 

j. Large notice boards and display boards 

k. Traffic signals and control boxes 

l. Vehicle detector pads 

m. Road drains or gullies 

n. Fire hydrants 

o. Stop valve and stand pipes 

p. Top of manholes (circular and square) 

q. Weigh bridge; and 

r. Services above ground (such as some water pipelines) 

6.6 BOUNDARY FEATURES 

a. Fences 

b. Gates 

c. Hedges 

d. Walls 

e. Burial grounds (indicate whether Muslim,. Chinese, Christian etc.) and 

f. Historical areas 

6.7 BRIDGE SITE 

a. Width of bridges 

b. Soffit levels of edge beam 

c. Carriage way 

d. Existing reserve 

e. Size, type and location of utility services adjacent and along the span of the bridge 

f. Spans and location of columns/piers 

g. Level of water and date taken 

6.8 RAILWAYS 

a. Railway running rails 

b. Points 

c. Bridges (over roads, river, etc.) 

d. Signal boxes 

e. Telephone points 

f. Telegraph poles, and 

g. Km posts (value to be noted) 

6-2 March 2009


6.9 SURVEY CONTROL 

a. Survey Department GPS and Boundary Marks for horizontal control 

b. Ground control points 

c. Permanent ground control markers 

d. Survey Department Bench Marks (BM) vertical control; and 

e. Temporary Bench Mark (TBM) established 

6.10 PLANTATIONS, TREES AND RECREATIONAL AREAS 

a. Playing fields 

b. Parks and open spaces 

c. Laid out pitches 

d. Prominent trees; and 

e. Land-use and vegetation etc. 

6.11 SLOPES AND EARTHWORKS 

a. Cutting and embankments 

b. Terraced slope 

c. Ornamental slopes 

d. Mounds 

e. Industrial waste; and 

f. Refuse tips 

6.12 WATER AND DRAINAGE 

a. Rivers (name to be indicated) 

b. Streams 

c. Water courses 

d. Ditches (width and depth to be indicated) 

e. Swamps 

f. Lined drains (type, size, depth to be indicated) 

g. Culverts with sizes and invert levels, including sketch of inlet and outlet structures such as wing 

wall 

h. Irrigation structures such as Weirs, bunds, spillways, barrage, floodgates, dams and floodwalls 

i. Pump station sites 

j. Tanks 

k. Sewer outfalls and top of manhole covers 

l. The top of all water features over 1.0 meter wide are to be detailed and the bottom of banks as 

indicated by the water level at the time of the survey. The direction of flow of all rivers, streams 

and watercourses is to be indicated 

m. Slopes with a height greater than 1.0 meter or too sharp a gradient to be shown by contours, 

including river banks, are to be shown by conventional markings and the top and bottom of 

slopes are to be shown as dotted lines; and 

n. Slope conventions are to be drawn as near as possible to indicate the actual shape of the slope 

face, i.e. all berms and terraces are to be detailed 

o. Flood spillways and closure bunds 

p. Tidal variation sites for tidal gate structures or bunds 

q. Highest known flood level 

Any other visible features not listed likely to affect design and later construction works are also to be 

shown. 

March 2009 6-3


REFERENCES 




[3] Digital Globe for Satellite Imagery at website http://www.digitalglobe.com 

[4] US Army Corps of Engineers website is accessible by keying in “us army corps of engineers 

hydrographic survey manual” then click “EM 1110-2-1003 Title: Engineering and Design – 

Hydrographic Survey” 





[7] GDM2000 Geodesy Section, Department of Survey and Mapping website 

http://geodesi.jupem.gov.my 

6-4 March 2009

Volume 6 – Geotechnical Manual, Site Investigation and Engineering ...

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